case study of geosynthetic materials

Geotextiles and Geomembranes 20 (2002) 343–365

A case study of geotextile-reinforcedembankment on soft ground

Dennes T. Bergadoa,*, Pham V. Longb, B.R. Srinivasa Murthyc

aSchool of Civil Engineering, Asian Institute of Technology, Bangkok 12120, ThailandbHydraulic Engineering Consultants No. 2, Ho Chi Minh City, Viet Nam

cCivil Engineering Department, Indian Institute of Science, Bangalore 560012, India

Received 30 March 2002; received in revised form 29 July 2002; accepted 18 August 2002

Abstract

Full-scale test embankments, with and without geotextile reinforcement, were constructed

on soft Bangkok clay. The performances of these embankments are evaluated and compared

with each other on the basis of field measurements and FEM analysis. The analyses of failure

mechanisms and the investigations on the embankment stability using undrained conditions

were also done to determine the critical embankment height and the corresponding geotextile

strain. The high-strength geotextile can reduce the plastic deformation in the underlying

foundation soil, increase the collapse height of the embankment on soft ground, and produce a

two-step failure mechanism. In this case study, the critical strain in the geotextile

corresponding to the primary failure of foundation soils may be taken as 2.5–3% irrespective

of the geotextile reinforcement stiffness.

r 2002 Elsevier Science Ltd. All rights reserved.

Keywords: Soft soil model; Mohr–Coulomb model; Critical height; Collapse height; Critical strain

1. Introduction

To investigate the performance of geotextile-reinforced embankment on softBangkok clay, three full-scale instrumented test embankments, with and withoutgeotextile reinforcements, were constructed and failures were achieved. Oneembankment was reinforced by multiple layers of low-strength, nonwoven, needle-punched geotextile (referred to as MGE embankment). The other was reinforced bya single layer of high-strength, composite nonwoven/woven geotextile referred to as

*Corresponding author. Tel.: +66-2-524-5512; fax. +66-2-524-6050.

E-mail address: [email protected] (D.T. Bergado).

0266-1144/02/$ - see front matter r 2002 Elsevier Science Ltd. All rights reserved.

PII: S 0 2 6 6 - 1 1 4 4 ( 0 2 ) 0 0 0 3 2 - 8

HGE embankment. An unreinforced embankment was also built nearby as a controlembankment referred to as CE embankment. The layout of the test embankments isgiven in Fig. 1 while the cross-sections are presented in Fig. 2. All embankments weredesigned with a side slope of 1.5:1. A canal of 2m deep and 7.5m wide was excavatedalong the toe of the embankments in order to reduce the amount of fill required forthe embankments to reach failure and to ensure that the failures occur in theintended direction. Ayutthaya silty sand, obtained from a local source, was usedas backfill material for the embankments. The instrumentation program, construc-tion procedure and monitored data for the CE and HGE embankments havebeen presented earlier by Bergado et al. (1994). The stability analyses of theseembankments were also presented by considering Bishop’s simplified method. Chaiet al. (1997) analyzed the performance of CE embankment by FEM using CRISPprogram. For the purpose of continuity, a brief of the full-scale load tests of the CEand HGE embankments are also presented in this paper, together with the details forMGE embankment. In this paper, the FEM undrained analyses are presented toinvestigate the failure mechanism for all the three embankments. The FEMconsolidation analyses results for CE and HGE embankments are compared with thefield monitored data. The FEM analysis have been performed using the softwarePLAXIS-version 6 (Vermeer and Brinkgreve, 1995).

2. Full-scale load tests

2.1. Geotextile reinforcement

The MGE embankment was reinforced by four layers of low-strength, nonwoven,needle-punched polypropylene geotextiles: one layer of Polyfelt TS700 and three

Fig. 1. Layout of the test embankments with locations of field investigation.

D.T. Bergado et al. / Geotextiles and Geomembranes 20 (2002) 343–365344

layers of Polyfelt TS420. The HGE embankment was reinforced by a single layer ofhigh-strength, nonwoven/woven geotextile (PEC200) placed directly on the naturalground surface. The properties of the geotextile are presented in Table 1.

2.2. Soil profile

The general soil profile and the basic soil properties are given in Fig. 3. Thetop 12m depth can be divided into four layers. The weathered crust consisting of

Fig. 2. Cross sections of the three test embankments.

Table 1

Properties of the geotextile reinforcements

Item Types of geotextile Nominal mass

(g/m2)

Secant

stiffness at

5.0% strain

(kN/m)

Ultimate

strength

(kN/m)

Remarks

1 Polyfelt TS 700 280 140 19 MGE

2 Polyfelt TS 420 130 70 9 MGE

3 Polyfelt PEC 200 700 1700 200 HGE

D.T. Bergado et al. / Geotextiles and Geomembranes 20 (2002) 343–365 345

heavily overconsolidated reddish-brown clay forms the uppermost 2m. This layeris underlain by soft, grayish clay down to 8.5m, followed by the thin 2.5mthick layer of medium stiff clay. A stiff clay layer is found underlying the mediumstiff clay layer with a sand layer at 14m depth. The shear strength measured from sixfield vane tests at depth intervals of 0.5m is also given in Fig. 3. The groundwaterlevel at the test site varied seasonally from a depth of 1.0–1.5m from the groundsurface.

2.3. Foundation instrumentation

The foundation instrumentation details of the MGE and HGE embankmentsare given in Fig. 4. Similar instrumentation arrangement was used for CEembankment but excluding the settlement plate (S2) and the standpipe piezometers(SP4, SP5, SP6). The total pore pressures were monitored at depths of 3, 5 and 7mfrom the ground surface. Twelve piezometers were also installed for eachembankment consisting of three hydraulic piezometers (HP1, HP2, HP3), sixopen standpipe piezometers (SP1 to SP6) and three pneumatic piezometers (PP1,PP2, PP3). In order to calculate the excess pore pressures, dummy piezometerswere also installed at the corresponding depths of 3, 5 and 7m at a distance ofmore than 20m away from the test embankment. Settlements at the groundsurface and subsurface locations were measured by means of surface and sub-surface settlement plates, respectively. The subsoil settlements were measuredat depths of 2, 4 and 6m from the ground surface. The lateral deflections in thesubsoils were measured by a digital inclinometer. One inclinometer was installed

Fig. 3. General soil profile and soil properties at the site.


vertically at the toe of each embankment down to the depth of 12m into the stiff claylayer.

2.4. Geotextile instrumentation

The displacement and strain in geotextile reinforcements were measured by wireextensometers, Glotzl extensometers and strain gages. The layout of geotextileinstrumentation in HGE embankment is shown in Fig. 5. Wire extensometers wereused to measure the total displacements of geotextile reinforcement. A total of 12wire extensometers (EG1 to EG12) were installed at 1m interval from 0 to 11mstarting from the center of embankment. Four Glotzl extensometers, G1–G4, wereinstalled (at 1m interval) starting at a distance of 3.0 and ending at a distance of6.0m from the center of embankment. Details of surface settlement plates,inclinometer and extensometer have been presented by Bergado et al. (1994).Special strain gages of the type EP-08-40 CBY-120 were used. These gages have a

nominal resistance of 120O and are 100mm long. At each of the measuring point,two strain gages were installed on both upper and lower sides of the geotextile, andwere connected in series. The strain gages were glued to the geotextile at both ends.The gages were then covered by a thin layer of rubber silicon for moisture protection.The strain gages were installed at four locations at the distances of 0.5, 1.5, 2.5 and

Fig. 4. Foundation instrumentation of the test embankments.


6.5m from the center of embankment corresponding to notations L1, L2, L3 and L4,respectively.

2.5. Embankment construction and behavior

Before construction of the test embankments, the ground surface was excavated toabout 0.20m depth and leveled. The canal with dimensions of 2m depth from theoriginal ground surface and 7.5m width at the bottom was excavated along theproposed failure side of all the embankments.

Fig. 5. Geotextile instrumentation in HGE embankment.


The embankment loading was calculated as the product of the fill thickness andthe total unit weight considering the effects of rainfall during construction. The netembankment height is defined as the difference between the current elevation ofembankment crest and the original elevation of the embankment base. The CEembankment was constructed in layers with compaction lift thickness of about0.33m. The dry density of 17 kN/m3 and average moisture content of 9% weremaintained. The total unit weight was 18.5 kN/m3. In all the three cases, very smalldeformations were observed at embankment heights lower than 2.5m. The rate ofsettlement and lateral movement increased significantly when the embankmentheights exceeded 3.5m. The CE embankment reached a net height of 4m (i.e. 4.32mfill thickness or 80 kPa embankment loading) 24 days after the start of embankmentconstruction. The corresponding maximum settlement and lateral movementmeasured at this time were 0.32 and 0.17m, respectively. No other constructionactivity was done and the embankment was stable until the morning of the 25th day.A crack of about 5mm wide and 3m long was observed in the morning of the 25thday along the bottom of the canal near the settlement plate, S6. The width of thecrack opened quickly to about 10mm wide, and then the embankment collapsedwithin 10min. The vertical crack and the collapse of embankment crest areillustrated in Fig. 6. The failure surface was estimated with the bamboo sticksembedded during construction, approximated as a circular surface.Subsequently, the MGE and HGE embankments were built at the same time and

at almost the same rate of filling. Construction procedure and quality control werealso kept the same as that of the CE embankment. When the two embankmentsreached the net height of 3.75m, one small crack of about 5mm wide was observed

Fig. 6. Cross section of MGE and CE embankment after collapse.


near the surface settlement plate S6 in the bottom of the canal. In contrast with thecase of the unreinforced embankment, the crack gradually developed to about10mm wide and 3m long the next day with no further development observed in thelater days. At this embankment height of 3.75m, the embankment loading was about75 kPa and the measured strains in geotextile were smaller than 1.0%. The rates ofsettlement, lateral displacement and geotextile strain increased significantly withfurther increase in height. When the embankments reached 4.2m net height, themeasured maximum settlements were 0.33 and 0.40m for HGE and MGEembankments, respectively. The corresponding maximum strains in the geotextilewere about 2.3% and 3%, and the most strained point in the geotextile is locatedabout 3.5m from the centerline of both embankments. The failure of MGEembankment that concurrently induced failure of HGE embankment occurred onthe 62nd day. All instruments of HGE embankment were still functioning except theinclinometer casing. The maximum strain in geotextile of HGE embankmentmeasured after the failure was 8.5%, which indicated that the failure has caused anadditional 6% strain in the geotextile. At this stage, the geotextile reinforcements inthe HGE embankment were not ruptured while in the MGE embankment they werecompletely ruptured. The HGE embankment construction was then continued to anet height of 6m at which stage it collapsed. The corresponding settlement measuredat the center of the embankment was 0.67m. The maximum strain in geotextile ofHGE embankment was in the order of 12–14%, which is the range of rupture strainfrom wide-width in-air tensile tests. A comparison of the failure planes of CE andreinforced embankments for both MGE and HGE embankments are presented inFigs. 6 and 7.

Fig. 7. Cross section of HGE and CE embankments after primary failure.


It is noted that there were 35 rainy days during the construction of MGE andHGE embankments. Consequently, the measured average field moisture content ofMGE and HGE fill was about 13% corresponding to the total unit weight of19.2 kN/m3, and was about 4% higher than that of the CE embankment with totalunit weight of 18.5 kN/m3. Therefore, the actual collapse loads of MGE and HGEembankments are 88 kPa (i.e., 4.2m net embankment height plus 0.4m maximumsettlement=4.6m fill thickness) and 128 kPa (i.e., 6m net embankment height plus0.67m maximum settlement=6.67m fill thickness), which are 1.1 and 1.6 timeshigher than that of the CE embankment (80 kPa and 4.32m fill thickness),respectively.

3. Finite element modeling

3.1. Finite element simulation

The finite element meshes are presented in Figs. 8 and 9 for CE and HGEembankments, respectively. Both consolidation and undrained analyses were carriedout. The consolidation analysis simulating the actual construction conditions wasconducted in order to consider the effects of consolidation process due to thedifference in construction sequences of the test embankments.The 6-noded triangle, linear-strain elements are employed for both foundation

soils and embankment fill. The geotextile reinforcement and soil-geotextile interfaceare simulated by 3-noded geotextile elements and 6-noded flat elements, respectively.The finite element mesh configuration is based on the actual soil profile (Fig. 3)which consisted of four layers: weathered crust from 0 to 2m, soft clay from 2 to8.5m, medium clay from 8.5 to 11m, and stiff clay from 11 to 14m. The soft clay isthen divided into three sublayers from 2 to 4m, 4 to 6m and 6 to 8.5m to better

Fig. 8. Finite element modeling for CE embankment.


simulate the depth effect. The piezometric drawdown due to excessive groundwaterpumping (Bergado et al., 1988) is considered in calculating the in situ stresses andpore pressures of the subsoils. The lower boundary with fixed displacements and freedrainage is set at 14m depth from the ground surface where the sand layer wasencountered.The embankment fill is modeled by three and five layers of elements for CE and

HGE embankments, respectively (Figs. 8 and 9). The canal excavation and theembankment construction are taken into account by removing or adding thecorresponding elements according to the construction sequence. In the PLAXISprogram, the load caused by the self-weight of removing or adding elements areautomatically applied by small increments in order to satisfy the tolerable errors(smaller than 5%) for both local and global equilibrium.

3.2. Material models and parameters

In the PLAXIS program, both modified cam-clay model (MCC) by Burland(1965) and the soft soil model (SSM) by Vermeer and Brinkgreve (1995) whichresembles the modified cam-clay model are available. However, for lightlyoverconsolidated soil (OCR o2), it has been suggested to use the SSM rather thanMCC because of the mesh sensitivity of the softening behavior in the latter modeland the improved behavior in one-dimensional compression of the SSM. Moreover,the Mohr–Coulomb failure criteria incorporated in SSM can represent the failurebehavior of the lightly overconsolidated soil having an apparent cohesion.Therefore, the SSM is used for both soft clay (2 to 6m depth) and medium stiffclay (6 to 8.5m depth). However, both MCC and SSM are not suitable for theheavily overconsolidated weathered crust. In this study, the elastic–perfectly plasticMohr–Coulomb model is adopted for this type of soil. For the stiff clay layer below11m depth, the elastic model is employed for simplicity since its stress state under

Fig. 9. Finite element modeling for HGE embankment.


embankment loading is far from failure. The model parameters of foundation soilsfor FEM analyses that are basically interpreted from the soil investigation data aregiven in Table 2.The SSM, for soft and medium clay layers, requires six parameters: friction angle,

f; cohesion c0; modified compression ratio, l�; modified swelling ratio, k�; at-restlateral earth pressure coefficient in normally consolidated state, K0ðNCÞ ¼ð1� sin f�Þ; and Poisson’s ratio, n0: The FEM prediction of embankment failureheight is very sensitive with the selected strength parameters, f0 and c0: The cohesiondepends on the current overburden, s0vo; the overconsolidation ratio, OCR, and thefriction angle, f0: In this study, the friction angle of the subsoils is adopted as 231based on extensive test results on soft Bangkok clays (Balasubramaniam, 1991). Thecohesion values are back-calculated from the failure of CE embankment in additionto deriving from the standard field vane shear tests by the following approximation:

c0 ¼ msuv �1þ 2K03

s0vo tan f0; ð1Þ

where suv is the average field vane strength, m is the correction factor, and K0 is theat-rest earth pressure coefficient, estimated as follows:

K0 ¼ K0ðNCÞOCRm; ð2Þ

In Eq. (2), the value of m ¼ 0:3 (Ladd, 1991) and the earth pressure coefficient innormally consolidated state, K0ðNCÞ ¼ 1� sin f

0; as given by Jaky (1948), are used.The correction factor, m; in Eq. (1), is obtained by matching the calculated andmeasured failure heights of CE embankment and a value of m of 0.77 was obtained.Then, the corresponding apparent cohesion, c0; can be obtained as presented inTable 2.The modified compression ratio, l�; is computed from the compression index, Cc;

which is obtained from the conventional oedometer test results. The empiricalcorrelation using the initial water content (Bergado et al., 1995a) has been used tofurther modify l� to account for the variation with depth. The value of k� is taken as0.2 times that of l� as suggested by Vermeer and Brinkgreve (1995). Poisson’s ratiofor a soft clay depends on the confining stress status level and hence a variable valuewith depth has been adopted.

Table 2

Selected parameter for subsoils in FEM analyses

Depth of

soil layers

Model c0 (kPa) F0 (1) l� K� n0 G kv � 10�4

(m/day)

OCR

0.2–2.0 M–C 20.0 23 0.33 1500 2.0 7.84

2.0–4.0 SSM 4.0 23 0.15 0.030 0.33 1.0 2.11

4.0–6.0 SSM 2.5 23 0.15 0.030 0.33 1.0 1.66

6.0–8.5 SSM 2.0 23 0.13 0.026 0.25 1.0 1.53

8.5–11.0 SSM 2.0 23 0.08 0.08 0.20 2.0 1.46

Note: M–C=perfectly elastic Mohr–Coulomb model; SSM = soft soil model.


For the heavily overconsolidated weathered crust, the elastic–perfectly plasticMohr–Coulomb model requires five parameters: friction angle, f0; cohesion, c0;dilatancy angle, c; shear modulus, G; and Poisson’s ratio, n0: The dilatancy angle hasbeen assumed to be zero for this soil. The strength parameters, f0 and c0; and theshear modulus, G; are obtained from large-scale laboratory test on weathered claycompacted to field conditions.For the embankment fill, the elastic–perfectly plastic Mohr–Coulomb model was

used. In this model, the shear modulus, G; can be expressed as a power function ofthe stress level (Vermeer and Brinkgreve, 1995) as follows:

G ¼ Grefp�

pref

� �m

; ð3Þ

where m is the power number, pref the reference pressure, taken as 100 kPa, Gref thereference shear modulus corresponding to p� ¼ pref ; where p� is calculated from theeffective mean stress, p0; as follows:

p� ¼ p0 þ c0 cot f0: ð4Þ

The apparent cohesion, c0; and the friction angle, f0; of embankment fill aredetermined from the large direct shear tests which were tested at the shearing rate of1mm/min that can be assumed as drained tests for the case of sand. The shearmodulus, G; and the power number, m; as well as the dilatancy angle, c, wereobtained from drained triaxial shear tests and also were confirmed from a FEM backanalyses of large size direct shear tests (Bergado et al., 1996). The selected parametersare given in Table 3.In the soil–geotextile interface, the elastic–perfectly plastic model is used to

simulate the constitutive relation. In the PLAXIS program, the compressionmodulus is related to the shear modulus assuming a fixed value of Poisson’s ratio of0.45. The shear modulus, Gi; and the strength parameters at the interface can beautomatically generated from that of the confining soil using the interactioncoefficient, Ri: The values of Ri obtained from large direct shear tests (Bergado et al.,1995b) were used in this analysis. The value of Ri was found to be 1.0 for all casesexcept TS 700 on weathered crust, which is equal 0.9.For geotextile reinforcement, the linear tension–strain relation was used in the

PLAXIS program for geotextile elements. The only property for this model is anaxial stiffness, S; defined as the product of Young’s modulus, E; and the cross-sectional area, A; of the fabric. The average stiffness of high-strength geotextilePEC200 is independent of the confining pressure and can be taken as 1700 kN/m for

Table 3

Selected parameters for embankment fill in FEM consolidation analysis

Embankments c0 (kPa) F0 (1) Gref (kPa) m n0 C0 (1) gt (kN/m3)

CE embankments 15 30 3065 0.71 0.33 8 18.5

HGE embankments 10 30 3065 0.71 0.33 8 19.2


the range of strain from 0% to 12%. The creep strains and the influence oftemperature on the in-soil stiffness of the reinforcements in the short period of theembankment construction were ignored.

4. FEM undrained analyses for embankment stability

The FEM undrained analyses for embankment stability were carried out for all thethree test embankments. These are parametric studies using the same properties ofembankment fill to verify the improvement of stability in the geotextile-reinforcedembankments and to further investigate the failure mechanisms and to directlycompare the stability behavior of the three test embankments. In FEM modeling ofMGE embankment, the three layers of geotextile TS420, which are spaced at onecompaction lift thickness of 0.33m, are simulated as one layer having an equivalentstiffness of the three layers (70� 3=210 kN/m).

4.1. Collapse and critical heights

From the case histories of HGE embankment and Sackville embankment (Roweet al., 1995; Rowe and Hinchberger, 1998), it appears that the height of a geotextile-reinforced embankment can be increased after the failure of foundation soil (primaryfailure). Therefore, it is necessary to distinguish the collapse height, Hf ; and thecritical height, Hc; of an embankment on soft ground. The collapse height (ofreinforced or unreinforced embankment) can be defined as the height at which anyattempt to increase the height of the fill will yield no increase in the net embankmentheight. In other words, the collapse height corresponds to the maximum netembankment height that can be constructed. Thus, the collapse height can beobtained directly from the plot of net embankment height versus embankmentsettlement. The net embankment height is defined as the difference between thecurrent elevation of the embankment crest and the original elevation of theembankment base. The critical height of a reinforced embankment corresponds tothe collapse height of an unreinforced embankment on the same foundation subsoil.Therefore, from Figs. 10a and b the net collapse heights of CE, MGE and HGEembankments would be 4.04, 4.65, and 5.96m, respectively, while the correspondingnet critical heights of MGE and HGE embankments are 4.40 and 4.80m.Furthermore, as observed from these figures, using higher geotextile stiffness, morereduction in foundation deformations and higher critical height can be achieved. Itshould be noted that the critical height and the collapse height of a reinforcedembankment defined herein can be considered as the heights corresponding to thefailure of foundation soil (primary failure) and the collapse of embankment and itsfoundation as a whole (secondary failure), respectively. For unreinforced CEembankment, the FEM results indicate that the critical height coincides with thecollapse height.


4.2. Critical strain in geotextile reinforcement

The critical strain, ec; is defined as the maximum strain mobilized in thereinforcement when the embankment reaches critical height, H ¼ Hc (i.e. primaryfailure). The FEM calculated maximum geotextile strains computed from FEManalyses in MGE and HGE embankments are plotted in Fig. 11 in relation with themaximum settlement and net embankment height. From this figure, it can be foundthat the critical strains in both embankments have almost the same magnitude ofabout 2.3% and this agreed well with the experimental critical height in the field. A

Fig. 10. FEM undrained analyses for the three test embankments. (a) Net embankment height versus

maximum settlement. (b) Net embankment height versus maximum lateral displacement.


comparison of critical strains measured for various geotextile-reinforced testembankments is presented in Table 4. The variables in the comparison are stiffnessof geotextile, soil properties and the embankment heights. It is interesting to notethat the critical strains in all the cases varies only in a narrow range of 2.5–3.0%.In order to verify the effects of geotextile stiffness and the subsoil condition on the

critical strain in geotextile, additional parametric FEM undrained analyses wereconducted for reinforced embankment on soft Bangkok clay at Nong Ngu Hao(NNH) site where the soft clay deposit is thicker and the rate of soil strength increasewith depth is smaller than those at the AIT site. The geotextile stiffness of 2000 and5000 kN/m were examined. The critical strains of 2.8% and 2.4% can be obtainedcorresponding to the reinforcement stiffness of 2000 and 5000 kN/m.Thus, it may be concluded that the subsoil profile and the geotextile stiffness have

very little effect on the critical strain. However, the difference between the collapse

Fig. 11. Net embankment height and geotextile strain versus settlement in FEM undrained analysis of

AIT test embankments.


height and the critical height is dependent on both geotextile stiffness and foundationsoil.

5. FEM consolidation analyses

The consolidation analyses, simulating the actual construction conditions, wereconducted for CE and HGE embankments in order to consider the effects of theconsolidation process associated with the difference in the construction sequencesand are compared with the measured data.

5.1. Responses of foundation soil

The typical profiles of lateral displacement under the toe of HGE and CEembankments are presented in Figs. 12 and 13, respectively. Generally, the FEMresults are comparable with the monitored data except for the prior failure case ofCE embankment, which may be due to excess plastic zone formation.Both FEM and measured data show the zone of highest lateral displacement

that occurred within the depth of 2–4m coinciding with the weakest soil layer. Thelateral displacement in an HGE embankment is substantially smaller than that of theCE embankment at embankment loading higher than 65 kPa. The lateraldisplacements were almost the same as the embankment loading increase up to65 kPa. Thus, the geotextile reinforcement mainly reduced the plastic lateraldisplacement. This is consistent with the interpretation of settlements. It should bementioned that the maximum lateral and vertical displacements in HGE embank-ment at 4.2m net height (q ¼ 88 kPa) are almost the same as that of CE embankmentprior to collapse at 4.0m height (q ¼ 80 kPa). This demonstrated that the foundationof the HGE embankment had failed at an embankment height of 4.2m. The failureof HGE embankment at this height, therefore, may also be considered as the primaryfailure.

Table 4

Comparison of critical strains for different test embankments

Item Test

embankments

Geotextile

stiffness

Critical strains

(%)

Embankment

critical height

(m)

Reference

1 Almere 2000 kN/m 2.5 2.05 Rowe (1992)

2 Sackville 1466 kN/m 3.0 5.7 Rowe et al.

(1995)

3 Guiche 2250 kN/m 3.0 7.0 Delmas et al.

(1992)

4 Present study 350 kN/m 2.3 4.4 Present study

1700 kN/m 2.3 4.8


In FEM analysis, the collapse load can be determined as the load that triggeredthe continuous lateral displacements. Consequently, the loading of 78 and 123 kPaare the FEM calculated collapse loads of CE and HGE embankments, respectively(Fig. 10a). The collapse loads are calculated based on the fill thickness, i.e., netembankment height plus maximum settlement of the embankment. The correspond-ing actual collapse loads of CE and HGE embankments were 80 and 128 kPa. Thus,good agreement between calculated and actual collapse loads has been obtained.

5.2. Displacements and strains in geotextile

The variation of geotextile strain versus time combined with the correspondingFEM results are presented in Fig. 14. Up to day 61, the calculated strains agreed wellwith the mean values of measured strains. Beyond this point the FEM results aresmaller than the measurements. This is because the induced failure on day 61resulted in an additional strain of 6%, which has not been included in the FEManalysis.

Fig. 12. Lateral displacement profiles in CE embankment.


Fig. 14. Geotextile strain versus time in HGE embankment.

Fig. 13. Lateral displacement profiles in HGE embankment.


5.3. Responses at soil-reinforcement interface

The shear stress distribution along the soil–geotextile interface obtained by FEManalysis are plotted in Fig. 15a and b corresponding to the net embankment heightsof 4.2m (q ¼ 88 kPa) and 6.0m (q ¼ 128 kPa) of the HGE embankment. At anembankment load of 88 kPa, which can be considered as the maximum service load,the interaction mechanism between the soil and geotextile is the direct shear mode, asmanifested by the opposing directions of the mobilized interface shear between theupper interface and the lower interface of the geotextile (Fig. 15a). However, prior tocollapse at q ¼ 128 kPa, the pullout mode dominated, as manifested by the same

Fig. 15. (a) Shear stresses at interfaces of PEC200/soil at Hnet ¼ 4:2m. (b) Shear stresses at interface ofPEC200/soil at Hnet ¼ 6:0m.


direction of the mobilized interface shear between the upper and lower interfaces ofthe geotextile reinforcement (Fig. 15b). Considering the magnitude of mobilizedshear stresses along the interfaces, it can be found that the interface shear stresses aremuch smaller than the shear strength of the surrounding soils even at collapse load.With the undrained shear strength of the weathered crust beneath the geotextile of37.5 kPa, the maximum value of interface shear stresses is less than one-half of theavailable shear strength. Therefore, there was no slippage between soil and geotextileand the pullout failure did not occur in this test embankment.

5.4. Failure mechanism

The contours of relative shear stress, defined as the ratio of shear stress to theavailable shear strength, are illustrated in Figs. 16a and b for CE embankmentcorresponding to embankment load of q ¼ 61 kPa and at failure (q ¼ qf ¼ 80 kPa),respectively. The darkest shade zones in these figures represent the zones with unityvalue of relative shear stress ratio i.e. the soil is considered as at plastic state orfailure. Fig. 16a indicates that the local failures appear in the soft clay layer at anembankment load of 61 kPa which is in accordance with the interpretation from themeasured pore pressures and deformation where the local failure load was found tobe 65 kPa. According to q ¼ 61 kPa, the value of qf =q is 1.3. Thus, a safety factoragainst bearing failure of foundation soil of 1.3, as used in conventional design, can

Fig. 16. (a) Contours of relative shear stress in CE embankment at q ¼ 61 kPa (qf =q ¼ 1:3). (b) Contoursof relative shear in CE embankment at failure q ¼ qf ¼ 80 kPa, contour interval=0.1, maximumvalue=1.0.


be considered as the limit to avoid the large deformation due to the development ofplastic zone in foundation soils. Moreover, the local failures that started at qf =q ¼1:3 as obtained herein also implied that the lower bound limit analyses, in which thefailure load is defined as the load corresponding to the failure at the first point in thesoil mass, underestimated the embankment stability. Increasing the embankmentloading to the failure load of qf ¼ 80 kPa, the plastic zones progressively developedto be a continuous band as seen in Fig. 16b. Thus, the failure surface has beenformed in the foundation soil while the shear strength in the embankment fill has notfully mobilized yet. However, even without the increase of embankment load, thedeformation in the embankment fill may continue because of the continuousdisplacement of foundation soils at plastic state together with the undrained creepdeformations at a high stress level. Consequently, the shear stress in embankment fillcontinued to mobilize until softening. Subsequently, the complete collapse of theembankment occurred. This explained the delay in the failure of CE embankmentthat occurred about 14 h after the failure load had been fully reached. Such failuremechanism suggested that the strength parameters used in stability analyses ofunreinforced embankment on soft Bangkok clay should be selected corresponding tolarge strain at critical state for both foundation soils and embankment fill.The progressive mobilization of relative shear stress of the HGE embankment are

reflected in Figs. 17a and b corresponding to embankment loads of q ¼ 88 and

Fig. 17. (a) Contours of relative shear stress in HGE embankment at q ¼ 88 kPa (Hnet ¼ 4:2m). (b)Contours of relative shear in HGE embankment at failure q ¼ 128 kPa, contour interval=0.1, maximumvalue=1.0.


128 kPa. At embankment loading of 88 kPa, the continuous band of plastic zone wasformed as seen in Fig. 17a. Increasing the embankment fill, the failure zonedeveloped further as reflected in Fig. 17b. Considering the failure zone in HGEembankment, it is confirmed that even without the failure of MGE embankment onday 61, the primary failure in HGE embankment might have also taken place at theembankment load around 88 kPa. Associated with the primary failure is themobilization of the localized geotextile strength that resulted in the increase ofembankment height until the embankment collapsed (secondary failure) accom-panied by the rupture of geotextile. This can be referred to as the two-step failuremechanism of high-strength geotextile-reinforced embankment on soft ground. Thisfailure mechanism is similar to the case history at Sackville embankment describedby Rowe et al. (1995). In this embankment, the primary failure was observed at 5.7mfill thickness with a sudden settlement of about 0.5m associated with the additionalstrain in geotextile of about 3%. Afterwards, the embankment fill was successfullyadded up to the embankment collapse at 8.5m fill thickness. The mobilized strain ingeotextile prior to primary failure was in the range of 2.3–3% in both HGE andSackville embankments, even though their geometry and foundation soil profile arenot the same. At this strain level, the improvement in embankment stability may notbe considerable. Nevertheless, both case histories indicated that the high-strengthgeotextile reinforcement can significantly increase the collapse height after theprimary failure of foundation soils. This is an important aspect that should beconsidered in practical applications.

6. Conclusions

The deformations and stability of geotextile-reinforced embankment duringconstruction up to failure of two test embankments have been successfullyinvestigated by finite element method using the PLAXIS program. From this study,the following conclusions and recommendations can be drawn:

(1) Both measured data and FEM results showed that the high-strength geotextilecan reduce significantly the plastic deformations in the underlying foundationsoils.

(2) The case histories and the results of theoretical analyses using FEM suggestedthe two-step failure mechanism of high-strength geotextile-reinforced embank-ment on soft ground. The primary failure of foundation soils may occur at thesame deformations as that of the unreinforced embankment just prior tocollapse. The secondary failure or collapse of embankment and foundation as awhole may occur with the rupture or pullout of the geotextile reinforcement.

(3) Depending on the stiffness of the geotextile reinforcement, the field measure-ment and the results of back-analysis indicated that the geotextile reinforcementcan increase the collapse height of embankment on a soft ground. In this casestudy, one layer of high-strength geotextile increased the collapse height up to1.5 times higher than that of the unreinforced case.


(4) The results of this study as well as other three case histories presented havesubstantiated that the critical strain, elc; in geotextile corresponding to theprimary failure of foundation soils is not much dependent on the geotextilestiffness and the subsoil conditions. In practical design, it can be taken as2.5–3%.

Acknowledgements

With deep appreciation and gratitude, the authors would like to acknowledge thefinancial and technical support from Polyfelt Geosythetics of Austria. Sincere thanksare due to Dr. P. Delmas for his technical assistance in the project.

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case study of geosynthetic materials

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