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I
SELF-CONSOLIDATING CONCRETE FOR PRECAST,
PRESTRESSED CONCRETE BRIDGE ELEMENTS
By Lama Baali
February 2009
Department of Civil Engineering and Applied Mechanics McGill University Montreal, Quebec
Canada
A thesis submitted to McGill University in partial fulfilment of the requirements of the degree of a Masters in Civil Engineering
The following thesis presents the results of four full scale beams tests as part of a research program
conducted at McGill University. The purpose is to study the applicability of existing design provisions, in
the American Association of State Highway and Transportation Officials (AASHTO) specifications, for
the use of self-consolidating concrete (SCC) in precast pretensioned bridge girders.
The test specimens had an overall length of 31 ft (9.4m) with a center-to-center span of 29 ft (8.8m). They
were cast in four batches with different concrete attributes: two non air-entrained SCC mixtures and two
high-performance concretes. For each type, compressive strengths of 8,000 and 10,000 psi (55.2 and 69
MPa) with release strengths of 5,000 and 6,250 psi (34.5 and 43 MPa) at 18 hours, respectively, were
tested. Each girder was prestressed with eight Grade 270 seven-wire low-relaxation prestressing strands
of 0.6 in (15.2 mm) diameter. Six of the strands were straight and two were harped twice, 4’-11” (1.5 m)
from mid-span. The specimens were supported on neoprene bearing pads at their ends, and were tested
with two equal point loads located 4’-11” (1.5 m) from mid-span.
This research project demonstrated that the shear failure of the girders exceeded the predicted nominal
shear resistance given by the 2004 AASHTO Specifications. The experimental flexural resistance also
exceeded the predicted nominal resistance.
II
Résumé
Le présent mémoire expose les résultats de quatre poutres pleine grandeur faisant partie intégrante d’une
étude effectuée à l’Université McGill. Le but de cette étude est de valider l’applicabilité des provisions de
conception existantes, de l’Association Américaine des Autoroutes d’État et des Officiers de Transport
(norme AASHTO), pour l’usage de béton autoplaçant (BAP) dans les poutres précontraintes et
préfabriquées de ponts.
Les spécimens testés ont une longueur maximale de 31 pieds (9.4 m) avec une distance du centre au
centre de 29 pieds (8.8 m). Les poutres ont été coulées une à la fois avec différentes sortes de béton: deux
d’entres-elles à partir de béton autoplaçant sans air entrappé, et deux avec du béton haute-performance.
Pour chaque sorte, une résistance compressive de 8,000 et 10,000 psi (55.2 et 69 MPa) avec une
résistance, avant de précontraindre le béton, de 5,000 et 6,250 psi (34.5 et 43 MPa) à 18 heures,
respectivement, ont été testées. Chaque poutre était précontrainte avec huit tendons, grade 270, de 0.6 in
(15.2 mm) de diamètre. Six de ces tendons étaient horizontaux alors que deux étaient inclinés 59 pouces
(1.5 m) de chaque bord de l’axe central. Les spécimens étaient supportés aux deux extrémités sur des pads
de néoprène et étaient testés avec deux charges concentriques situées 59 pouces (1.5 m) de l’axe central.
Cette recherche à démontrer que la capacité en cisaillement des poutres testées excédait les valeurs
nominales prévues par les normes AASHTO 2004. Les valeurs expérimentales de la résistance à la
flexion des poutres aussi excèdent les valeurs nominales prédises.
III
Acknowledgments
The author would like to express deep gratitude to Professor Denis Mitchell for the supervision, guidance
and support throughout this thesis. The author would also like to extend his gratitude to Dr. William Cook
for his significant help during the tests, his patience and his valuable explanations during the writing of
this thesis.
The great help of the technical staff of the Jamieson Structure Laboratory was also very much
appreciated. The assistance of Mareck Przykorski, Ron Sheppard, John Bartczak and Damon Kiperchuck
is greatly appreciated for the successful completion of the experiments. The help of Patrick Moubarak,
Dean MacDougall, Lesley Wake, and Yahya Baali in the laboratory work is deeply appreciated.
Finally, the author would like to thank her parents, Yasser and Hana, her sisters and brothers, Dania,
Roula, Ammar and Yahya, and her brother-in-law, Assad, for always being there for her and always
encouraging her to reach her goals.
Lama Baali, 2009
IV
Table of Contents
Abstract .......................................................................................................................................................... I Résumé ......................................................................................................................................................... II Acknowledgments ....................................................................................................................................... III Table of Contents ........................................................................................................................................ IV
List of Figures ........................................................................................................................................... VII List of Tables .............................................................................................................................................. IX
Chapter 1: Introduction and Literature review .............................................................................................. 1 1.1 Chapter overview ................................................................................................................................ 1
1.5.2 Testing fresh-state properties of SCC ........................................................................................ 40
1.5.2.1 The Slump flow and T-20 tests ........................................................................................... 41
1.5.2.2 The J-ring test ...................................................................................................................... 42
1.5.2.3 The L-box and filling ability tests ....................................................................................... 42
1.5.2.4 The column segregation test................................................................................................ 43
1.5.2.5 The surface settlement test .................................................................................................. 44
1.5.2.6 The rheology test ................................................................................................................. 44
1.5.3 Comparison between SCC and normal concrete ........................................................................ 46
1.6 Objectives of this Research Program ................................................................................................ 47
Chapter 2: Experimental Program ............................................................................................................... 48 2.1 Design of the beam specimens .......................................................................................................... 48
2.2 Instrumentation and Test Setup ......................................................................................................... 54
2.2.3.1 Transfer length measurements ............................................................................................ 61
2.2.3.2 Strand set measurements ..................................................................................................... 64
2.3 Material properties ............................................................................................................................ 65
2.3.1 Concrete material properties ...................................................................................................... 65
2.3.1.1 Mix proportioning of HPC and SCC used for girder casting .............................................. 65
2.3.1.2 Concrete’s mechanical properties field testing program ..................................................... 66
2.3.1.3 Fresh properties of HPC and SCC used for girder casting .................................................. 70
2.3.1.4 Semi-adiabatic temperature measurements ......................................................................... 72
VI
2.3.1.5 Mechanical properties of deck slab concrete ...................................................................... 73
2.3.2 Reinforcing steel and prestressing steel properties .................................................................... 73
2.4 Test procedure ................................................................................................................................... 75
Figure 1: Stress distribution in a reinforced concrete beam containing flexural cracks ................. 3 Figure 2: Design of transverse reinforcement for shear: compression field theory ........................ 6 Figure 3: Average strains in web elements ..................................................................................... 7 Figure 4: Design of transverse reinforcement for shear: modified compression field theory ........ 8 Figure 5: Ideal stress-strain diagram for prestressing steel ........................................................... 30 Figure 6: Prefabrication procedure ............................................................................................... 36 Figure 7: Flame cut unstressed strands at bulkhead ...................................................................... 37 Figure 8: Typical atmospheric steam curing cycle ....................................................................... 38 Figure 9: Slump flow test .............................................................................................................. 41 Figure 10: J-ring test ..................................................................................................................... 42 Figure 11: L-box test ..................................................................................................................... 43 Figure 12: Column segregation test .............................................................................................. 44 Figure 13: Rheology test ............................................................................................................... 46 Figure 14: Section properties of AASHTO type II girders .......................................................... 49 Figure 15 - Details of precast pretensioned AASHTO II girders ................................................. 49 Figure 16: Details of the pretensionning ....................................................................................... 50 Figure 17: Details of non-prestressed reinforcement .................................................................... 52 Figure 18: Details of reinforcing bars in end region ..................................................................... 53 Figure 19: Details of the cross section .......................................................................................... 53 Figure 20: Details of stirrups and interface shear reinforcement .................................................. 54 Figure 21: Jacking of 0.6 in (15.2 mm) diameter strands ............................................................. 55 Figure 22: Reinforcing cage in formwork prior to casting ........................................................... 56 Figure 23: HPC girder casting operation ...................................................................................... 57 Figure 24: SCC girder casting operation ...................................................................................... 57 Figure 25: Steam curing chamber used for AASHTO Type II girders ......................................... 58 Figure 26: Temperature history of chamber and concrete during steam curing – HPC 8,000 psi
(55.2 MPa) ................................................................................................................ 59 Figure 27: Temperature history of chamber and concrete during steam curing – SCC 8,000 psi
(55.2 MPa) ................................................................................................................ 59 Figure 28: Temperature history of chamber and concrete during steam curing – HPC 10,000 psi
(69 MPa) ................................................................................................................... 60 Figure 29: Temperature history of chamber and concrete during steam curing – SCC 10,000psi
(69 MPa) ................................................................................................................... 60 Figure 30: Flame cutting of strands during prestress release ........................................................ 61 Figure 31: Transfer length measuring device ............................................................................... 61 Figure 32: Locations of transfer length strain gauges ................................................................... 62 Figure 33: Measured strains used to estimate transfer length ....................................................... 63 Figure 34: Measurement of strand set ........................................................................................... 64 Figure 35: Temperature rise of concrete under semi-adiabatic conditions ................................... 72 Figure 36: Typical stress-strain relationships for reinforcing No. 3 and No. 5 bars ..................... 74 Figure 37: Typical stress-strain relationship for 0.6 in (15.2 mm) diameter strand ..................... 75 Figure 38: Test setup and locations of LVDTs ............................................................................. 77 Figure 39: Specimen before testing .............................................................................................. 78
VIII
Figure 40: Variation of camber with time ..................................................................................... 79 Figure 41: Moment versus central deflection – Specimen H8 ...................................................... 80 Figure 42: Moment versus central deflection – Specimen S8 ...................................................... 81 Figure 43: Moment versus central deflection – Specimen H10 .................................................... 81 Figure 44: Moment versus central deflection – Specimen S10 .................................................... 82 Figure 45: Moment versus longitudinal strain response at the level of the straight strands ......... 83 Figure 46: Shear versus deflection at load point – Specimen H8 ................................................. 84 Figure 47: Shear versus deflection at load point – Specimen S8 .................................................. 84 Figure 48: Shear versus deflection at load point – Specimen H10 ............................................... 85 Figure 49: Shear versus deflection at load point – Specimen S10 ................................................ 85 Figure 50: Variation in the average vertical strains obtained from the LVDT rosette readings on
the end that failed in shear ........................................................................................ 87 Figure 51: Specimen H8 just before (a) and right after (b) failure ............................................... 88 Figure 52: Specimen S8 just before (a) and right after (b) failure ................................................ 89 Figure 53: Specimen H10 just before (a) and right after (b) failure ............................................. 90 Figure 54: Specimen S10 just before (a) and right after (b) failure .............................................. 91 Figure 55: Comparison of moment versus central deflection responses for the four specimens . 94 Figure 56: Comparison of shear versus deflection at loading point responses for the four
Table 1: ACI 318-08 Design of shear reinforcement .................................................................... 16 Table 2: ACI 318-08 Particular considerations for High Strength Concrete ................................ 17 Table 3: Value of θ and β for sections with transverse reinforcement ...................................... 19 Table 4: Value of θ and β for sections with less than minimum transverse reinforcement ....... 20 Table 5: AASHTO LRFD Shear design reinforcement ................................................................ 21 Table 6: Standard prestressing strands, wires, and bars ................................................................ 31 Table 7: Requirements for prestressing tendons specified by ASTM .......................................... 31 Table 8: Specimen Identification .................................................................................................. 48 Table 9: Summary of 95% AMS transfer lengths ......................................................................... 63 Table 10: Average values of strand set ......................................................................................... 64 Table 11: Mixture proportioning of SCC and HPC for girders .................................................... 65 Table 12: Concrete Compressive Strength of SCC and HPC mixes used for girders ................. 67 Table 13: Elastic Modulus of SCC and HPC mixes used for girders ........................................... 68 Table 14: Flexural Strength of SCC and HPC mixes used for girders ......................................... 69 Table 15: Fresh properties of SCC and HPC mixes used for girders ........................................... 70 Table 16: Concrete temperature under semi-adiabatic conditions ................................................ 72 Table 17: Mixture proportioning and fresh properties of deck slab concrete for all four girders 73 Table 18: Fresh properties of the deck slab concrete .................................................................... 73 Table 19: Average mechanical properties of reinforcing bars ...................................................... 74 Table 20: Typical mechanical properties of prestressing strand ................................................... 75 Table 21: Flexural responses of the specimens ............................................................................. 82 Table 22: Shear responses of the specimens ................................................................................. 86 Table 23: Comparison of flexural responses of four specimens ................................................... 93 Table 24: Comparison of shear responses of four specimens ....................................................... 95 Table 25: Tensile strengths at 56 days .......................................................................................... 95
1
Chapter 1: Introduction and Literature review
The purpose of this thesis is to study the applicability of existing design provisions, in the
American Association of State Highway and Transportation Officials (AASHTO) specifications,
for the use of self-consolidating concrete (SCC) in precast pretensioned bridge girders. The
research program consisted of the construction and testing of four full-scale precast, prestressed
bridge girders with selected Self-Consolidating Concrete (SCC) mixtures and companion high
performance concrete (HPC) specimens. This thesis provides both SI Units and U.S. Customary
Units due to the fact that the research was carried out for the U.S. National Cooperative Highway
Research Program (NCHRP).
1.1 Chapter overview
This chapter presents a review of the literature regarding the various topics that will be studied in
this thesis.
The literature review will begin with a brief history describing the developments in understanding
of shear behaviour and the relevant codes of practice (see Section 1.2).
Section 1.3 will summarize the benefits of prestressing, and will discuss the codes and standards
used for design.
Section 1.4 describes the precasting operations: history, benefits and tradeoffs.
Section 1.5 introduces the concept of Self-Consolidating Concrete (SCC).
Finally, Section 1.6 presents the different objectives of the research program.
1.2 Shear design
This report reviews the literature on the shear problem in reinforced and prestressed concrete
beams, with particular attention devoted to beams constructed with HPC. Section 1.2.1 reviews
the historical development of the research; Section 1.2.2 presents the 318-08 ACI Code (ACI 318
2008); and Section 1.2.3 covers the 2004 AASHTO LRFD (AASHTO 2004) bridge design
specifications. Section 1.2.4 summarizes a study comparing the ACI code approach to that used in
the AASHTO LRFD specifications.
2
1.2.1 History behind the shear design
In the early ages of reinforced concrete studies, pioneers developed two mechanisms for
estimating shear failures in reinforced concrete members. The first mechanism considered
horizontal shear as the basic cause of shear failures (ACI-ASCE Committee 326, 1962). This
seemed a reasonable approach at the time when engineers and researchers were familiar with the
action of web rivets in steel girders and shear-keys in wooden beams, for which shearing stresses
were computed using the classical equation:
IbVQ
=υ Eq. 1
Where:
=υ Unit horizontal shear stress at a distance y from the neutral axis
V = Total vertical shear at the section
Q = First moment of the part of the cross-sectional area cut off at distance y from the neutral
axis, with respect to the neutral axis
I = Moment of inertia of the cross-sectional area with respect to the neutral axis
b = Width of the cross section at a distance y from the neutral axis
Reinforced concrete beams were tested as an extension of the older materials assuming that the
concrete could only resist low horizontal shearing stresses, and that vertical stirrups acted as
shear-keys for higher shearing stresses.
The second mechanism considered diagonal tension as the basic cause of shear failures (ACI-
ASCE Committee 326, 1962). In 1899, W. Ritter presented a clear explanation of the diagonal
tension using a 45º truss model (Collins and Mitchell, 1997). He stated that stirrups resisted
tension not horizontal shear, and suggested that the design of stirrups for vertical shear be
determined from the following expression:
sjdfAV vv=
Eq. 2
Where:
vA = Total cross-sectional area of one stirrup
vf = Allowable stress in the stirrups
3
jd = Internal moment arm
s = Spacing of stirrups in the direction of the axis of the member
Ritter’s design expression for vertical stirrups is identical to that appearing in modern design
specifications of most countries.
Discussions between the proponents of horizontal shear and diagonal tension continued for nearly
a decade until laboratory tests resolved the issue mainly through the efforts of E. Mörsch in
Germany. He pointed out that, if a state of pure shear stress exists, then a tensile stress of equal
magnitude must exist on a 45-degree plane (ACI-ASCE Committee 326, 1962). Mörsch
explained Ritter’s model in more detail. He also predicted that the shear stress would reach its
maximum value at the neutral axis and would remain constant from the neutral axis down to the
flexural steel (Figure 1). The value of the maximum shear stress was evaluated by:
bjdV
=υ Eq. 3
Where:
V = Shear in beam
b = Width of rectangular section
d = Effective depth to center of gravity or reinforcement
j = Ratio of lever arm of tensile reinforcement to effective depth to steel computed by
straight-line theory for ordinary reinforced concrete beams
Figure 1: Stress distribution in a reinforced concrete beam containing flexural cracks [Adapted from Collins and Mitchell, 1997)]
4
Succeeding papers by Mörsch in 1906 and 1907 (ACI-ASCE Committee 326, 1962) explained
the diagonal tension mechanism and listed the following arguments against the horizontal shear
concept:
1- The ultimate minimal shearing stresses in beams without web reinforcement, as computed by
Eq. 3, are close to the tensile strength of concrete. Punching tests, on the other hand, indicate that
the shearing strength of concrete is considerably greater than its tensile strength. Hence, shear
failure in beams is due to tension, not horizontal shear (ACI-ASCE Committee 326, 1962).
2- The effectiveness of stirrups far surpasses the values computed by the horizontal shear theory.
The effectiveness of stirrups derived from the tensile force transmitted across a diagonal tension
crack is in better accord with tests (ACI-ASCE Committee 326, 1962).
3- Eq. 3, which expresses the nominal shearing stress, is intended to be only a nominal measure
of diagonal tension (ACI-ASCE Committee 326, 1962).
By 1910, a return to Ritter’s pioneering concepts had been made, though the concepts of
horizontal shear kept on reappearing in literature up until the early 1960s.
In the 1950s, researchers such as Zwoyer and Siess (1954), Bresler and Pister (1958), Guralnick
(1959), and Walther (1962) studied the stress conditions in the concrete above flexural cracks in
order to develop expressions for the shear capacity of members containing flexural-shear cracks.
They typically assumed that all of the shear would be carried in the flexural compression zone
and hence believed that the actual shear stress distribution was significantly different from that
shown in Figure 1. The uncertainty about the actual distribution of shear stresses over the section
caused engineers to refer to Eq. 3 as the “nominal” shear stress (Collins and Mitchell, 1997).
In 1963, ACI Committee 318 pointed out that this classical computation of the shear stress
involved an oversimplified concept of diagonal tension stress. Since the actual distribution of the
shear stress had not yet been fully clarified, the use of an average shear stress seemed advisable.
The ACI 318-63 Code (ACI 318, 1963) found the use of the moment arm, jd, unwarranted. The
ultimate shear stress was altered to average stress on the full effective cross section, and became:
bdV
=υ Eq. 4
5
Where b, previously defined as the “width of rectangular section”, was reduced for I and T-
sections to the width of the web. If the web was to be slightly tapered, an average web width was
to be used in computations.
In addition, Ritter’s equation (Eq. 2) for the design of vertical stirrups was adjusted to:
sdfAV vv=
Eq. 5
The 1963 ACI code also incorporated two empirical equations for nominal shear stress at the
flexure-shear cracking load. Both of these equations were developed by the ACI/ASCE Shear
Committee by reviewing available research. The equations were simplified to ease every-day
design work, and placed as such that the ultimate strength of beams be governed by flexure
failures rather than by shear failures.
The first expression for the nominal shear stress at the flexure-shear cracking load of a reinforced
concrete beam is:
dbfdbMVdfV wcwwc
'' 5.325009.1 ≤⎟⎠⎞
⎜⎝⎛ += ρ (inch-pound system) Eq. 6
dbfdbMVdfV wcwwc
'' 29.01716.0 ≤⎟⎠⎞
⎜⎝⎛ += ρ (mm-Newton system)
Where:
V = External shear at diagonal tension cracking of the section considered
V/M = Ratio of shear to moment at section considered
wρ = Ratio of non-prestressed tension reinforcement = bdAs
The second equation is a simplified version of the first one:
dbfV wc'2= (inch-pound system) Eq. 7
dbfV wc'17.0= (mm-Newton system)
These two equations, valid for members subjected to shear and flexure only, are found in the ACI
318-08 Code.
In 1973, the ACI-ASCE Shear Committee wrote “During the next decade it is hoped that the
design regulations for shear strength can be integrated, simplified and given a physical
6
significance so that designers can approach unusual design problems in a rational manner” (ACI-
ASCE Committee 426, 1973). Nonetheless, in 1984, MacGregor described the shear equations in
the ACI code as “empirical mumbo-jumbo” (MacGregor, 1984).
In 1974, a “rational” solution for the design of reinforced concrete beams to resist torsion was
developed by Mitchell and Collins (1974) known as the “compression field theory”. This concept
followed the strain compatibility conditions in the “tension field theory” developed by Wagner
(1929) to describe the post-buckling behaviour of thin webs of steel girders. He assumed that
after buckling, the thin webs would not resist compression and that the shear would be carried by
a field of diagonal tension. The compression field theory is similar to Neilsen’s (Brook and
Brown, 1967) lower bound solution: it described shear behaviour through the entire cracked range
up to failure. The compression field theory idealized the diagonally cracked concrete as a material
with coinciding principal stress and strain axes which develop to satisfy both equilibrium and
compatibility of strains.
Figure 2: Design of transverse reinforcement for shear: compression field theory
[Adapted from Collins and Mitchell, 1997]
The vertical component of the diagonal compressive force in the concrete, which is inclined at the
angle θ to the longitudinal axis, must equal the applied shear force (Figure 2) and hence:
⎟⎠⎞
⎜⎝⎛ +⎟⎟⎠
⎞⎜⎜⎝
⎛=
θθ
tan1tan2 jdb
Vfw
Eq. 8
In turn, the diagonal compression in the concrete transfers vertical forces to the stirrups so that:
θtanjdV
sfA vv =
Eq. 9
7
The longitudinal component of the diagonal compression in the concrete is equilibrated by
tension:
θtanVfAfAN ppxxv =+=
Eq. 10
The compression field theory also assumed that concrete, once cracked, carries no tension and
that the shear is carried by a field of diagonal compression. Applying Wagner’s approach to
reinforced concrete resulted in the following expression for the angle of inclination of the
diagonal compression (Figure 3):
2
22tanεεεε
θ−+
=t
x Eq. 11
Figure 3: Average strains in web elements [Adapted from Collins and Mitchell, 1986]
In 1982, Vecchio and Collins tested reinforced concrete panels under biaxial stresses and pure
shear. They found that the principal compressive stress in the concrete, 2f , is a function not only
of the principal compressive strain 2ε , but also of the coexisting principal tensile strain 1ε . In
fact, the concrete web is not only in compression in direction 2, but is also in tension in direction
1. They suggested the following parabolic stress-strain relationship:
⎥⎥⎦
⎤
⎢⎢⎣
⎡⎟⎟⎠
⎞⎜⎜⎝
⎛−⎟⎟
⎠
⎞⎜⎜⎝
⎛=
2
'2
'2
max,22 2cc
ffεε
εε
Eq. 12
Where:
0.11708.01
1'
max,2 ≤+
=εcf
f
Eq. 13
By using these equations, it was possible to predict not only the strength but the load-deformation
response of members loaded in shear. However, it was found that this theory overestimates the
8
deformations and underestimates the strengths because it neglects the contribution of the tensile
stresses in cracked concrete (Collins and Mitchell, 1997). By 1986, Vecchio and Collins came up
with the “modified compression field theory”. This theory stated that shear force is resisted by the
diagonal compressive stresses, 2f , together with the diagonal tensile stresses, 1f , accounting for
the contribution of the concrete in tension, even after it has cracked (Figure 4).
Figure 4: Design of transverse reinforcement for shear: modified compression field theory [Adapted from Collins and Mitchell, 1997]
From Mohr’s stress circle, the following expression for 2f was derived:
12 tan1tan ff −⎟
⎠⎞
⎜⎝⎛ +=
θθυ
Eq. 14
Where:
jdbV
w
=υ
Compared with the previous equilibrium Eq. 8 of the compression field theory, Eq. 14
incorporates the concrete tensile stresses contributing to carrying the load. The diagonal
compressive stresses 2f push apart the flanges of the beam while the diagonal tensile stresses 1f
pull them together. The vertical imbalance is carried by tension in the web reinforcement. The
equilibrium requirement is expressed as:
( ) wvv bff
sfA
θθ 21
22 cossin −=
Eq. 15
Combining Eq. 14 and 15 quantified the concrete contribution, a value up until then always
approximated:
9
θθ cotcot 1 jdbfjdsfA
V wvv +=
Eq. 16
cs VVV +=
V = Steel contribution + Concrete contribution
Eq. 17
In fact, between 1904 and 1922, several hundred reinforced concrete beams were tested by Talbot
at the University of Illinois (Hognestad, 1952) and by Moritz at the University of Wisconsin
(Hognestad, 1952). These tests demonstrated that the stirrup stresses were considerably lower
than those predicted by the 45º truss model. This was due to the neglection of tensile stresses in
the concrete and the choice of 45º for the compressive strut inclination. According to the 45º truss
model, a beam without any transverse reinforcement would have zero shear strength. In order to
account for the contribution of the tensile stresses in the concrete, the first ACI code, in 1910
(ACI 1910) stated:
“In calculating web reinforcement the concrete shall be considered to carry 40 psi (0.275
MPa), the remainder to be provided for by means of web-reinforcement in tension.”
As the typical concrete compressive strength in those days was 2,000 psi (13.8 MPa), this
working stress code permitted concrete tensile stress at working load of '02.0 cf , equating about
'04.0 cf at ultimate.
In 1989, the ACI Code (ACI 318, 1989) separated the shear resistance of a beam into two
components, the “concrete contribution” due to tensile stresses in the concrete and the “steel
contribution” due to tensile stresses in the stirrups.
A number of experiments were conducted to study the influence of concrete strength on the shear
strength of reinforced concrete beams. Mphonde and Frantz (1985) tested 12 reinforced concrete
beams with 'cf ranging from 3,500 to 13,000 psi (24.2 to 89.7 MPa). They concluded that the
ACI Code (Eq. 6, 16 and 17), are conservative for all values of 'cf . Soon after, Elzanati, Nilson
and State (1986) and Nilson (1987) tested 9 reinforced concrete beams with
psifc 500,9' > (65.55 MPa) for shear strength and compared the results with 6 beams of
psifc 800,5' ≤ (40 MPa). They concluded that the ACI Code Equation 6 is unconservative by 10
to 30% for beams combining high strength concrete with medium to high shear span ratios and
typical or relatively low longitudinal steel ratios.
10
In 1989, Johnson and Ramirez (1989) tested 8 rectangular beams with concrete strength ranging
from 5,000 to 10,500 psi (34.5 to 72.5 MPa), and with web reinforcement ratios sv of 0 to 100
psi (0.69 MPa). Their results indicated that the overall reserve shear strength after diagonal
cracking, cfail VV − , diminishes as the concrete compressive strength increases, for a constant
reinforcement ratio. This data was used to justify the ACI 318-89 Code expressions in ∫ 1.2.1.11
which limits 'cf to 10,000 psi (69 MPa) or increases the minimum amount of reinforcement by a
factor equal to 'cf /5,000 psi (34.5 MPa) but less than 3 times the amount provided for concrete
with 'cf < 5,000 psi (34.5 MPa).
Roller and Russell (1990) reviewed 150 tests and confirmed the validity of the ACI Code
equations for shear. They also tested 10 beams with 'cf ranging from 10,500 psi to 18,000 psi
(72.5 to 124.2 MPa), all of which confirmed the findings of Johnson and Ramirez’s research.
Throughout the years, the provision requiring a sudden increase in the minimum amount of
transverse reinforcement for concrete strengths between 10,000 and 15,000 psi (69 and 103.5
MPa) was replaced by a gradual increase in the minimum vA , as 'cf increases.
Despite all the tests done in the past, research is still to be continued. In fact, due to the number of
variables involved, a general shear theory has been evasive. Design has been based on empirical
evidence, resulting in almost as many empirical equations as investigators. This basis has
provided a multitude of design equations for the design of structures in shear. For instance, the
ACI Building Code provides five different equations to evaluate the concrete contribution to
shear resistance for nonprestressed members, and three for prestressed members. To calculate cV
according to the AASHTO design specifications is dependent on the version of specifications
used. In general, the 16th edition conforms to the ACI Building Code. However, the AASHTO
LRFD bridge design specifications have introduced substantially different provisions for shear
design, based on the modified compression field theory.
1.2.2 ACI 318 Building Code (31808)
ACI 318, while generally providing ease in calculation, has been identified as having many
shortcomings including lack of conservatism for lightly reinforced cross-sections, for sections
utilizing high strength concrete, and deep sections.
11
The ACI 318-08 code starts by stipulating that the design shear strength of a member, nVφ , must
be greater that the factored shear, uV :
nu VV φ≤ ACI 11- 1
Where:
=φ Strength reduction factor, taken equal to 0.85 for shear
With the nominal shear strength nV given as:
=nV sc VV + ACI 11- 2
Where:
=cV Nominal shear strength provided by concrete
=sV Nominal shear strength provided by steel
Assuming that all stirrups yield at failure, the shear resisted by the stirrups perpendicular to the
axis is computed by:
sdfA
V ytvs =
ACI 11- 15
Where:
=vA Area of shear reinforcement within spacing s, 2in )( 2mm
=ytf Specified yield strength of transverse reinforcement, psi )(MPa
=sd Number of vertical stirrups spaced s apart, in a beam of d effective depth, crossed by a
45º crack
=d Distance from extreme compression fiber to centroid of the prestressed and
nonprestressed longitudinal tension reinforcement, if any, but need not be taken less than
0.80h
12
1.2.2.1 Shear strength provided by concrete for nonprestressed members
The ACI code assumes that cV is equal to the shear strength of a beam without stirrups, which in
turn is taken equal to the load at which inclined cracking occurs. For members subjected to
flexure and shear only:
dbfV wcc'2λ= , lb ACI 11- 3
dbfV wcc'17.0= , N
This equation was developed by ACI-ASCE Committee 326 in 1962 who also permitted a more
detailed calculation for cV . For members subjected to flexure and shear only:
dbfdbM
dVfV wcwu
uwcc
'' 5.325009.1 ≤⎟⎟⎠
⎞⎜⎜⎝
⎛+= ρλ , lb ACI 11- 5
dbfdbM
dVfV wcw
u
uwcc
'' 29.01716.0 ≤⎟⎟⎠
⎞⎜⎜⎝
⎛+= ρ , N
Where:
=wb Web width, in (mm)
='cf Concrete compressive strength, psi (MPa)
=uM Factored moment at section, lbin ⋅ )( mmN ⋅
=wρ Longitudinal reinforcement ratio
=λ Modification factor reflecting the reduced mechanical properties of lightweight concrete,
all relative to normalweight concrete of the same compressive strength
1= for normalweight concrete
=dV
M
u
u Shear span to depth ratio, a/d
0.1≤u
u
MdV
13
For members subjected to axial compression and axial tension, equations ACI 11-3 and ACI 11-5
are not applicable. The code provides a simplified and a more detailed equation for each of these
loading cases.
For members subjected to axial compression, the simplified method sets cV as:
dbfA
NV wcg
uc
'
200012 λ⎟
⎟⎠
⎞⎜⎜⎝
⎛+= , lb ACI 11- 4
dbfA
NV wc
g
uc
'
14117.0 ⎟
⎟⎠
⎞⎜⎜⎝
⎛+= , N
Where:
=uN Factored axial force normal to cross section; to be taken as positive for compression and
negative for tension, lb (N)
0> in this case
=gA Gross area of concrete section, 2in ( )2mm
Whereas the detailed method, for members subjected to axial compression, specifies:
g
uwcw
m
uwcc A
Ndbfdb
MdV
fV500
15.325009.1 '' +≤⎟⎟⎠
⎞⎜⎜⎝
⎛+= λρλ , lb ACI 11- 6
g
uwcw
m
uwcc A
Ndbfdb
MdV
fV29.0
129.01716.0 '' +≤⎟⎟⎠
⎞⎜⎜⎝
⎛+= ρ , N
Where:
=mM Factored moment modified to account for effect of axial compression
( )8
4 dhNM uu−
−= , lbin ⋅ )( mmN ⋅ ACI 11- 7
=u
u
MdV
Not restricted to 1.0
14
For members subjected to axial tension, Clause 11.2.1.3 for the simplified method stipulates that
cV is taken as zero, unless a more detailed analysis is performed using:
0500
12 ' ≥⎟⎟⎠
⎞⎜⎜⎝
⎛+= dbf
ANV wc
g
uc λ , lb ACI 11- 8
029.0
117.0 ' ≥⎟⎟⎠
⎞⎜⎜⎝
⎛+= dbf
AN
V wcg
uc , N
Where:
=uN Factored axial force normal to cross section; to be taken as positive for compression and
negative for tension, lb (N)
0< in this case
1.2.2.2 Shear strength provided by concrete for prestressed members
The situation for members with an effective prestress force of at least 40 percent of the tensile
strength of flexural reinforcement is different and is managed by other equations for cV .
The simplified method sets cV as:
dbM
dVfV w
u
pucc ⎟⎟
⎠
⎞⎜⎜⎝
⎛+= 7006.0 'λ , lb ACI 11- 9
dbf wc'2λ≥
dbf wc'5λ≤
dbM
dVfV w
u
pucc ⎟⎟
⎠
⎞⎜⎜⎝
⎛+= 8.405.0 ' , N
dbf wc'17.0≥
dbf wc'42.0≤
Where:
0.1≤u
pu
MdV
15
A more detailed method allows cV to be taken as the lesser of ciV and cwV .
ciV represents the nominal shear strength provided by concrete when diagonal cracking results
from combined shear and moment, and is calculated as:
dbfM
MVVdbfV wc
creidpwcci
'
max
' 7.16.0 λλ ≥++=, lb ACI 11- 10
dbfM
MVVdbfV wc
creidpwcci
'
max
' 14.005.0 ≥++= , N
Where:
=creM Moment causing flexural cracking at section due to externally applied loads
⎟⎠⎞⎜
⎝⎛ −+⎟⎠⎞⎜
⎝⎛
dpect
fffyI '6λ , lbin ⋅ ACI 11- 11
⎟⎠⎞⎜
⎝⎛ −+⎟⎠⎞⎜
⎝⎛
dpect
fffyI '5.0 , )( mmN ⋅
maxM and iV : Taken from the load combination causing maximum factored moment
cwV , on the other hand, represents the nominal shear strength provided by concrete when diagonal
cracking results from high principal tensile stress in the web
( ) ppwpcccw VdbffV ++= 3.05.3 'λ , lb ACI 11- 12
ppwpcccw VdbffV +⎟⎠⎞⎜
⎝⎛ += 3.029.0 ' , N
Where:
hVp 80.0≥
cwV : Computed as the shear force corresponding to dead load plus live load that results
in a principal tensile stress of psif c'4λ ⎟
⎠⎞⎜
⎝⎛ MPafc
'33.0 at the centroidal
axis of member, or at the intersection of flange and web when the centroidal axis
is in the flange
16
1.2.2.3 Minimum shear reinforcement
The shear design reinforcement, for prestressed and nonprestressed members, is summarized in
Table 7: Requirements for prestressing tendons specified by ASTM (ASTM A416, A421, A722)
Tendon Type Minimum Tensile
Strength, ksi (MPa)
Minimum Yield Strength, ksi
(MPa)
Minimum Elongation at
Rupture
% Gage Length, in (cm)
0.5 and 0.6 in. (12.7 and 15.24 mm) stress-relieved strand
270 (1862)
230 (1586) 3.5 24 (61)
0.5 and 0.6 in. (12.7 and 15.24 mm) low-relaxation strand
270 (1862)
245 (1689) 3.5 24 (61)
0.276 in. (7 mm) wire 235 (1620)
200 (1379) 4.0 10 (25.4)
1, 1 1/4, and 1 3/8 in. (25.4, 31.75 and 34.9 mm) deformed prestressing bar
150 (1034)
120 (827) 4.0 20 db*
* bd : Nominal diameter of reinforcing bar
32
1.3.5.3 ACI 318‐08 maximum permissible stresses in concrete and reinforcement
Definition:
=pyf Specified yield strength of prestressing tendons, psi
=yf Specified yield strength of nonprestressed reinforcement, psi
=puf Specified tensile strength of prestressing tendons, psi
='cf Specified compressive strength of concrete, psi
='cif Compressive strength of concrete at time of initial prestress, psi
Stresses in concrete immediately after prestress transfer, before time-dependent prestress losses,
shall not exceed the following:
a) Extreme fiber stress in compression '60.0 cif
ACI clause 18.4.1 b) Extreme fiber stress in tension except as permitted in c) '3 cif c) Extreme fiber stress in tension at ends of simply supported members
'6 cif
Where computed tensile stresses exceed these values, bonded auxiliary reinforcement
(nonprestressed or prestressed) shall be provided in the tensile zone to resist the total tensile force
in concrete computed under the assumption of an uncracked section.
Stresses in concrete at service loads (after allowance for all prestress losses) shall not exceed the
following:
a) Extreme fiber stress in compression due to prestress plus sustained load, where sustained dead load and live load are a large part of the total service load
'45.0 cf
ACI clause 18.4.2 b) Extreme fiber stress in compression due to prestress plus total
load, if the live load is transient '60.0 cf
c) Extreme fiber stress in tension in precompressed tensile zone '6 cf
ACI clause 18.4.1
d) Extreme fiber stress in tension in precompression tensile zone of members (except two-way slab systems), where analysis based on transformed cracked section and on bilinear moment0deflection relationships shows that immediate and long-time deflections comply with the ACI definition requirements and minimum concrete cover requirements
'12 cf
33
Tensile stress in prestressing tendons shall not exceed the following:
a) Due to tendon jacking force But not greater than the lesser of puf80.0 and the maximum value recommended by the manufacturer of prestressing tendons or anchorages.
pyf94.0
ACI clause 18.5.1 b) Immediately after prestress transfer But not greater than puf74.0
pyf82.0
c) Post-tensioning tendons, at anchorages and couplers, immediately after tendon anchorage
pyf70.0
1.3.5.4 2004 AASHTO maximum permissible stresses in concrete and reinforcement
Concrete stresses before creep and shrinkage losses
a) Stress in compression of pre-tensioned members '60.0 cif
AASHTO clause 5.9.5.3 and 5.9.5.4
(Adapted: units in psi)
b) Stress in compression of post-tensioned members '55.0 cif c) In tension area with no bonded reinforcement 200 psi or
'3 cif d) Where the calculated tensile stress exceeds c), bonded
reinforcement shall be provided to resist the total tension force in the concrete computed on the assumption of an uncracked section. The maximum tensile stress shall not exceed:
'5.7 cif
Concrete stresses at service load after losses
a) Stress in compression '40.0 cf AASHTO clause 5.9.4.2
(Adapted: units in
psi)
b) Tension in the precompressed tensile zone - For members with bonded reinforcement '6 cf
For sever corrosive exposure conditions, such as coastal areas
'3 cf
- For members without bonded reinforcement 0
Cracking stresses: modulus of rupture rf from tests or if not available
For normal weight concrete '5.7 cf AASHTO clause 5.4.2.6
(Adapted: units in psi)
For sand-lightweight concrete '3.6 cf For all other lightweight concrete '5.5 cf
34
Prestressing steel stresses
a) Due to tendon jacking pupy ff 80.094.0 ≤ AASHTO clause
5.9.5.4 (Adapted: units in psi)
b) Immediately after prestress transfer pupy ff 74.082.0 ≤
c) Post-tensioning tendons at anchorage, immediately after tendon anchorage
pyf70.0
Where:
For low-relaxation: pupy ff 90.0=
35
1.4 Precasting
The following sections will look into the benefits of precasting, also known as prefabrication
(Section 1.4.1), the general method used in precast shops (Section 1.4.2), and the few techniques
used to make this practice efficient (Section 1.4.3). Section 1.4.4 presents a research conducted by
Mokhtarzadeh and French (2000) comparing different HPC attributes and mixtures.
1.4.1 Benefits of precasting
In 2001, the AASHTO Technology Implementation Group chose prefabricated bridge elements
and systems as one of the innovative technologies that promised the highest payoff. The FHWA
(Federal Highway Administration) – through its Innovative Bridge Research and Construction
program and the Resource Center – also championed prefabrication for accelerated construction.
Their vision was to solve bridge deterioration with accelerated construction through increased
prefabrication. As a matter of fact, prefabrication technology carries many advantages for bridge
owners, engineers, builders, and the traveling public (Naito, Brunn, Parent and Tate, 2005).
In the case of precast construction there are savings in formwork assembly, concrete casting, and
curing offsite in the controlled environment of a precast plant. In addition to the more rapid
construction on site, the quality of the components is improved, translating into lower longer
service life and lower life-cycle costs. Shipment of precast components to the job site rather than
cast-in-place construction also reduces transportation costs and the impact on the environment
(Naito, Brunn, Parent and Tate, 2005).
From a design point of view, fewer constraints, such as extreme elevations, long stretches over
water, and tight urban work zones, are to be considered and overcome. Safety is also improved
because of the reduced time exposure for workers and public who travel through construction
zones. Moreover, prefabricating takes elements and systems out of the critical path of a project
schedule (Naito, Brunn, Parent and Tate, 2005).
All factors put together, prefabrication makes some bridge design, whether involving a new
construction or rehabilitation, more feasible, affordable and constructible.
1.4.2 Production methods
Precasting plants follow a conventional method of fabrication. In order to be productive and
competitive, a cycle should not take more than 24 hours to be completed. It involves setting up
36
the formwork, the reinforcement and prestressing, casting, curing, releasing the prestressing once
the concrete has hardened and removing the formwork.
The following section represents the main steps employed at Schuylkill Products Inc., a certified
PCI plant in Cressona, Pennsylvania (Naito, Brunn, Parent and Tate, 2005).
The beams are normally cast on long stressing beds (Figure 6a). The casting bed is a multi strand
tensioning operation which allows for simultaneous stressing of all strands in the cross-section
(Figure 6b). This configuration also allows for slow release of prestress after casting operations
are complete, thus minimizing uneven release and unintentional damage to the precast members.
a) Stressing bed
b) Stressing strands
Figure 6: Prefabrication procedure (Naito, Brunn, Parent and Tate, 2005)
The reinforcement is then put in place and the formwork is closed. Concrete is placed and left to
cure overnight. Special techniques are put in place (Section 1.4.3) to accelerate this phase and
allow faster prestressing release, ideally the following morning, if the target strength is reached.
37
Figure 7: Flame cut unstressed strands at bulkhead
(Naito, Brunn, Parent and Tate, 2005)
The release of the prestressing is done by flame cutting (Figure 7). The formwork is then to be re-
oiled and prepared for a new production cycle.
1.4.3 Particularities of precast concrete
Successful precast operations are dependent on rapid gain of concrete strength, leading to a faster
production rate and a more beneficial financial outcome. In order to be able to release the stress
faster, steam curing is widely employed in precast prestressed concrete manufacture to accelerate
gain of strength (Benaim, 2008). A great deal of attention has been directed at the effect of this
accelerated curing on long-term ultimate compressive strength, durability, shrinkage and creep,
loss of prestress, etc. Other studies have been directed to the determination of the optimum cycle
for the steam-curing process.
Two methods of steam curing are used: live steam at atmospheric pressure used for enclosed cast-
in-place structures and large precast concrete units, and high-pressure steam in autoclaves, for
small manufactured units (Cement Association of Canada, 2006).
Steam curing at atmospheric pressure is generally done in an enclosed environment to minimize
moisture and heat losses (Cement Association of Canada, 2006). A typical cycle consists of
(Figure 8):
(1) An initial 3 to 5 hours delay prior to steaming, allowing for some concrete hardening
resulting in higher early strengths
(2) A period for increasing the temperature
38
(3) A period holding the maximum temperature constant, and
(4) A period for decreasing the temperature
Figure 8: Typical atmospheric steam curing cycle [Adapted from Cement Association of Canada, 2006]
Properly applied steam curing improves the quality of the concrete product. In fact, it reduces
drying shrinkage and creep when compared to regularly cured concrete at 73°C for 28 days
(Klieger, 1960, and Tepponen and Eriksson, 1987)
Steam curing at atmospheric pressure has been one of the most important techniques which have
made it possible to attain economical production of prestressed concrete elements by permitting
daily turnover of forms. It has also made it practicable to have a short cycle between manufacture
and erection, eliminating in large part the need for stockpiles and inventory.
1.4.4 Alireza Mokhtarzadeh and Catherine French’s experiment (2000)
Mokhtarzadeh and French (2000) examined the mechanical properties (compressive strength,
modulus of elasticity and tensile strength) of high performance concrete. They tested over 6,300
39
specimens from 142 HPC mixtures with 28-day compressive strengths in range of 8,000 to
18,600 psi (55.2 to 128.2 MPa). The material variables included the following:
- Composition of cementitious material: ASTM C 150 portlant cement and combinations of
ASTM C 68 Class C fly ash and silica fume different percentages
- Type and brand of cement (Types I and III from two different brands)
- Type of silica fume: dry-densified and slurry
- Type and brand of superplasticizer: five types
- Type and source of coarse aggregates: high and low-absorption limestone, granites, round
and crushed river gravel
- Aggregate gradation
- Maximum aggregate size: ½ and ¾ in (12.7 and 19 mm)
For each specimen cured in standard lime-saturated water, a companion specimen was heat-cured
in a monitored electronically-controlled environmental chamber. After casting, the curing routine
was: 3 hours at room temperature followed by a 2.5 hours period to increase the temperature from
120°F to 160°F (49°C to 71°C); the temperature was then to be kept constant over 12 hours to
finally return to room temperature over a 2 hours period.
Heat-cured specimens yielded higher early-age compressive strengths than moist-cured
specimens. At later ages, however, continuous application of moisture resulted in the continued
increase in compressive strength of moist-cured specimens, whereas the strength of the heat-cured
specimens levelled off.
Mokhtarzadeh and French also showed that moist-curing was essential for getting full advantage
of using fly ash in a mixture. In the absence of adequate moisture, any benefit from inclusion of
fly ash was limited to grain refinement of the cement matrix. Moist-cured specimens also proved
to have an increasing modulus of elasticity with time. In fact, at 182 and 365 days of age, moist-
cured cylinders had, 106 and 108% of the 28-day compressive strength value, whereas the heat-
cured were only at 94 and 96%, respectively.
The compressive strength developed at early ages was higher for moist-cured concrete made with
rapid hardening portlant cement (Type III) than for ordinary portlant cement (Type I).
40
1.5 Selfconsolidating concrete
Concrete placement in precast plants can be time consuming and lead to unsatisfactory results if
vibrations are not done properly. One innovative solution may lie in the use of a highly flowable
self-consolidating concrete (SCC) mix.
1.5.1 SCC characteristics
SCC is a non-segregating concrete that can flow and fill formwork without any mechanical
vibration. This highly flowable concrete was developed in Japan in the1980s as a solution to
improve the constructability of reinforced concrete structures (Ozyildirim and Lane, 2003). Since
no mechanical vibration is needed when placing this concrete, significant savings in labour costs
and construction time can be achieved. SCC also has the potential of increasing durability and
quality resulting in a lower cost of construction. Further advantages include noise reduction
during construction and a reduction of surface defects leading to a more appealing architectural
finish (Gurjar, 2004).
A typical SCC mix is designed by ensuring a proper flowability and viscosity in the fresh state
(Ozyildirim and Lane, 2003). Flowability is normally achieved by using high-range water-
reducing admixtures (HRWRA) while viscosity is ensured by using a proper selection of fines
and aggregates and by using a viscosity modifying admixture (VMA). The flowability of the
product allows for placement in members with high amount of reinforcement congestion. The use
of conventional concretes in this situation would require significant mechanical vibration and
could result in a risk of honeycomb formation.
On the other hand, self-consolidating concrete also has the potential of lowering the permeability
of the concrete resulting in reduced durability and an increase of the life-cycle costs.
1.5.2 Testing freshstate properties of SCC
Several methods can be used to evaluate the various properties of SCC in the fresh state. Tests
can be broadly split into two categories: free flow tests and restricted flow tests. These procedures
enable an assessment of the filling ability, passing ability, and segregation resistance of SCC.
Among the most common tests for assessing the free deformability of SCC is the slump flow test.
Methods that are typically used to assess the restricted deformability include the L-box and the J-
ring tests. Other tests, such as column segregation, surface settlement and rheology tests are
covered in the following section.
41
1.5.2.1 The Slump flow and T‐20 tests
A conventional slump cone is filled with concrete and placed on a Plexiglas table on which a
concentric diameter of 20 inches (50.8 cm) is marked (Figure 9). The cone is then lifted and the
time for the concrete diameter to reach the 20 in benchmark is recorded. This time is referred to
as the T-20 value and typically varies between 2 to 10 seconds for SCC. A higher T-20 value
indicates a more viscous mix: a concrete more appropriate in congested reinforcement. A lower
T-20 value may be appropriate for concrete that has to travel long horizontal distances without
much obstruction. When the concrete stops flowing, its final diameter (D-final) is measured
(Gurjar, 2004).
Figure 9: Slump flow test (Gurjar, 2004)
While performing the slump flow test, the resistance to segregation is observed through a Visual
Stability Index (VSI). The VSI is established based on whether bleed water is observed at the
leading edge of the spreading concrete of if aggregates pile at the center. VSI test ranks the
stability on a scale from 0, for “highly stable”, to 3, for unacceptable stability (Gurjar, 2004).
ASTM C 1611 provides descriptions of the surface bleed, mortar halo and aggregate distribution
to properly select the appropriate VSI.
Due to the simple nature of this test procedure, slump flow is one of the most common methods
used in practice to measure the workability of SCC in its fresh state.
42
1.5.2.2 The J‐ring test
The J-ring test is used to determine the passing ability of the concrete. It consists in a steel ring of
12 in (30.5 cm) diameter, drilled vertically with holes to accept threaded sections of
reinforcement bar, 4 in (10.2 cm) height (Figure 10). These sections of bar can be of different
diameters and spaced at different intervals (Gurjar, 2004).
Figure 10: J-ring test [Adapted from Gurjar, 2004]
The J-ring is placed centrally on the base-plate with the slump cone. The cone is filled without
tamping and lifted to allow the concrete to flow out freely (Gurjar, 2004). When the concrete
comes to rest, the final diameter is measured. The difference between the slump flow’s final
diameter and the J-ring flow’s final diameter is a measure of the passing ability. A difference of
less than 1 in (2.54 cm) indicates good passing ability. A difference greater than 2 in (5.1 cm)
indicates poor passing ability (ASTM C1621). The J-ring test, used in conjunction with the slump
flow also indicates the flowing ability of concrete.
1.5.2.3 The L‐box and filling ability tests
The L-box test measures different properties, such as flowability, blocking and segregation. The
vertical part of the box, with the extra adapter mounted, is first filled with concrete and left to rest
for a minute (Figure 11). The sliding gate is then lifted allowing the concrete to flow out of the
vertical part into the horizontal part of the L-box. On its way, concrete passes through a layer of
reinforcement with bars usually spaced 1.5 in (3.8 cm) apart. After the sliding gate is removed,
the time for the leading edge of the concrete to reach the 8 in and 16 in (20.3 cm and 40.6 cm)
mark is recorded (Gurjar, 2004).
43
Figure 11: L-box test [Adapted from Gurjar, 2004]
The concrete is then left to rest in the apparatus and the heights of the concrete at the end of the
horizontal portion, H2, and in the vertical section, H1, are measured. The blocking ratio, 21 HH ,
is known as the filling capacity. The filling capacity test evaluated both the narrow-opening
passing ability and the self-leveling ability simultaneously (Hwang, Khayat and Bonneau, 2006).
For most tests, the blocking ratio should range between 0.80 and 0.85 (Gurjar, 2004). A smaller
value would indicate that the concrete built a plateau behind the reinforcement layer by either
blocking or segregating. Blocking usually displays itself by coarse aggregates gathered between
the reinforcement bars. If coarser aggregates are distributed on the concrete surface all the way to
the end of the horizontal part, the concrete can be regarded as stable. Both blocking and stability
can be detected visually (Gurjar, 2004).
While the test does give valuable information about filling and passing ability, and to a lesser
extent, segregation resistance, the test is not as simple as the slump flow test. Since there are no
standardized dimensions, results from different test apparatuses cannot be compared directly
(Koehler and Fowler, 2003).
1.5.2.4 The column segregation test
The column segregation test evaluates the static stability of a concrete mixture by quantifying the
aggregate segregation. A column is filled with concrete and left undisturbed until the concrete
comes to a rest (Figure 12). The column is then separated into three or four pieces. Each section is
44
removed individually and the concrete from that section is washed over a No. 4 sieve. The
retained aggregate are then washed and weighed. A non-segregating mix will have a consistent
aggregate mass distribution in each section, as opposed to a segregating mix which will have
higher concentrations of aggregate in the lower sections.
The percent of static segregation, known as the column segregation index, is then evaluated
Variations of the concrete temperature under semi-adiabatic conditions are presented in Figure 35
and summarized in Table 16.
Figure 35: Temperature rise of concrete under semi-adiabatic conditions Table 16: Concrete temperature under semi-adiabatic conditions
H8 S8 H10 S10 Peak temperature oF (oC) 139.6 (60) 144.6 (62.5) 147.1 (64) 150.7 (66) Elapsed time hours 15.5 16.8 15.0 16.9
It can be noted that the two 10,000 psi (69 MPa) concrete mixtures, H10 and S10, developed
higher peak temperatures than both 8,000 psi (55.2 MPa) concrete mixtures, H8 and S8. They
however took longer times to reach those peaks. Moreover, for a given compressive strength, the
SCC mixtures reached temperatures similar but higher than their companion HPC mixtures.
Consequently, the SCC mixtures had longer elapsed times to reach peak temperatures compared
to their companion HPC mixtures. This can be due to the higher HRWRA dosage.
H8
H10
S8
S10
73
2.3.1.5 Mechanical properties of deck slab concrete
All four deck slabs were to have the same concrete properties. They were to simulate a cast-in-
place deck slab with conventional air-entrained concrete of regular compressive strength of 5,000
psi (34.5 MPa), with Type I cement.
Table 17 summarizes the concrete proportions used in the production of the ready-mix concrete
delivered to the structural laboratory of McGill University from Unibéton’s plant, while
Table 18 presents the fresh properties of each batch of concrete, as two were required.
Table 17: Mixture proportioning and fresh properties of deck slab concrete for all four girders Mixture Deck slab Compressive strength psi (MPa) 5,000 (34.5) w/cm 0.42 Cement Type I 3ydlb ( )3mkg 639 (380)
Water 3ydlb ( )3mkg 269 (160)
Sand 3ydlb ( )3mkg 1189 (707)
Coarse aggregate 3ydlb ( )3mkg 1822 (1083)
Retarder 3ydoz ( )3mml 17.7 (684)
Air entraining agent 3ydoz ( )3mml 8.8 (342)
Table 18: Fresh properties of the deck slab concrete Deck Slab H8 and S8 H10 and S10 Initial concrete temperature oF (oC) 77.9 (25.5) 80.6 (27) Slump in (mm) 3.74 (95) 3.74 (95) Air content % 6.1 7.6 Average compressive strength at 28 days psi (MPa) 5030 (34.7) 4690 (32.3) Elastic modulus at 28 days psi (MPa) 4590 (31.6) 4280 (29.5) Average compressive strength at 38 days psi (MPa) 5250 (36.2) 5090 (35.1)
2.3.2 Reinforcing steel and prestressing steel properties
2.3.2.1 Reinforcing steel
The girders contained two different sizes of reinforcing bars: No. 3, for the stirrups and interface
shear reinforcement and No. 5, for the longitudinal bars. Figure 36 shows the typical stress-strain
74
relationships for each one of these non-prestressed reinforcing bars. Their mechanical properties,
such as the average values of the yield stress, yf , the ultimate stress, uf , the strain at strain
hardening, shε and the ultimate strain, uε , are summarized in Table 19.
Figure 36: Typical stress-strain relationships for reinforcing No. 3 and No. 5 bars (Khayat and Mitchell, 2009)
Table 19: Average mechanical properties of reinforcing bars
nV kips (kN) 113 (153) 115 (156) 118 (160) 119 (161) It is noted that the experimentally determined maximum shears are considerably higher than the
predicted nominal shear resistance. There are several reasons for the conservative predictions:
• The tensile strengths obtained from the flexural beam tests indicate that the corresponding
code values are very conservative for the HPC and SCC concretes. In fact, clause 5.4.2.6 of
the 2004 AASHTO evaluates the tensile strength for normalweight concrete as '24.0 cf ,
where 'cf is in ksi. Table 25 allows a comparison between the experimental and predicted
values of the concrete at 56 days cured with the same conditions as the girders.
Table 25: Tensile strengths at 56 days Steam cured for 18 hours
+ Air-cured with the same conditions
as the girders
H8 S8 H10 S10
Concrete compressive strength
psi(MPa)
7,378 (50.87)
8,605 (59.33)
9,050 (62.40)
9,555 (65.88)
Experimental tensile strength
psi (MPa)
960 (6.62)
922 (6.36)
850 (5.86)
855 (5.9)
Theoretical tensile strength psi (MPa)
662 (4.56)
704 (4.85)
722 (4.98)
742 (5.12)
Ratio 1.45 1.31 1.18 1.15
96
As noted, the AASHTO underestimates the resisting capacity of the concrete in tension,
therefore lowering the predicted values.
• During the testing it was evident that the strength and stiffness of the top and bottom flanges
of the specimens considerably increased the shear strengths, with shear cracks of width 0.28
inch (7.1 mm) observed in the webs before shear failure occurred. In fact, looking at the
LVDT readings shown in Figure 50 demonstrate that the specimens were able to withstand
loads even after the stirrups had yielded: a point at which the girder is expected to fail.
Cracks were widening in the web but the concrete in the thick bottom flange had not yet
fully cracked.
Comparing the nominal shear resistance, nV , to the yield resistance, yV , shows predictions
that are also are met.
4.2.2 Comparison between the shear behaviour of SCC and HPC concrete
When compared to their companion girders, the SCC girders were found to have a smaller shear
force at the first shear cracking, crV ; a bigger shear force at the first stirrups yielding, yV ; but a
smaller maximum shear capacity, maxV . SCC girders also deflected less under the loading point.
97
Figure 56: Comparison of shear versus deflection at loading point responses for the four specimens
98
Chapter 5: Conclusions
The following conclusions arise from the construction and testing of the precast pretensioned
girders:
• The placement of the SCC mixture was successful: with casting only from one location at
mid-span of the girders, the concrete made its way to both ends without leaving any voids
even in areas of congested reinforcing steel. Moreover, there were fewer “bug holes” in the
SCC concrete than the HPC.
• There was no visible segregation of the concrete in all four specimens.
• During the steam-curing operation, the target chamber temperatures of 131 oF (55 oC) and 140
oF (60 oC) for both 8,000 and both 10,000 psi (55.2 and 69 MPa) concrete mixture,
respectively, was not achieved. The maximum concrete temperature reached, however, was
below the maximum temperature allowed of 150 oF (65.6 oC).
• The target compressive strengths at 18-hours, required for the prestress release, of both SCC
mixtures were met. The HPC mixtures showed values below but were very close to their target
strengths.
• The transfer lengths found for all four concrete types and mixtures were similar. Moreover,
they were all considerably shorter than values predicted by the ACI 318-08 Code and the 2004
AASHTO Specifications.
• The cracking moments were similar for the SCC girders and their companion HPC girders.
The uncracked and cracked stiffnesses for all four girders were similar, although the SCC
girders experienced slightly lower maximum moments.
• The HPC girders experienced flexural resistances which exceeded the predicted nominal
resistance using the 2004 AASHTO Specifications. For the 8,000 and 10,000 psi (55.2 and 69
MPa) concrete girders, the experimental flexural resistances exceeded the predicted
resistances by 3.8% and 8.3%, respectively.
• The SCC girders also experienced flexural resistances which exceeded the predicted nominal
resistance using the 2004 AASHTO Specifications. For the 8,000 and 10,000 psi (55.2 and 69
MPa) concrete girders, the experimental flexural resistances exceeded the predicted ones by
1.9% and by less than 0.1%, respectively.
99
• All four girders failed in shear after developing a significant number of wide shear cracks. The
maximum shear crack widths before failure were greater than 0.24 inch (6.1 mm). The stirrups
developed significant strains, beyond strain hardening, with the stirrups rupturing at failure.
• The cracking and maximum shears were similar for all four girders, although the SCC girders
experienced slightly lower resistances.
• The HPC and SCC girders experienced failure shears which exceeded the predicted nominal
resistance using the 2004 AASHTO Specifications. The experimental shear failures exceeded
the predicted values by 56% to 69%. This increased shear resistance was probably due to the
strength and stiffness of the top and bottom flanges of the AASHTO girders.
• The HPC girders experienced higher deflections and hence higher ductilities than their
companion SCC girders.
• The lower shear resistance and lower ductility experienced by the SCC girders is probably due
to the lower volume of coarse aggregate, which reduces aggregate interlock and results in a
lower energy absorption capability on the sliding shear failure plane.
100
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