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SEISMIC PERFORMANCE OF DEEP COLUMN-TO-BEAM WELDEDREDUCED BEAM SECTION MOMENT CONNECTIONS

J.M. Ricles, Lehigh University, U.S.A.X. Zhang, Lehigh University, U.S.A.

J.W. Fisher, Lehigh University, U.S.AL.W. Lu, Lehigh University, U.S.A

ABSTRACT

An experimental study was conducted to investigate the seismic behavior ofreduced beam section (RBS) moment connections to a deep wide flangecolumn. The test matrix for the experimental program consisted of six full-scale

interior RBS connections, where the column for the specimens ranged in depthfrom a W24 to a W36 wide flange section. All but one of the specimens had acomposite floor slab. The results from the study show that a composite floorslab provides restraint to the top flange of the beams; reducing the magnitudeof beam top and bottom flange lateral movement in the RBS, column twist, and

strength degradation due to beam instability in the RBS. The performance ofeach of the test specimens was found to meet the seismic connectionqualification criteria in Appendix S of the AISC Seismic Provisions, and therebyhave sufficient ductility for seismic resistant design. The results of the

experimental study, along with a nonlinear finite element study were used todevelop seismic design recommendations for RBS connections to deep

columns.

INTRODUCTION

RBS beam-to-column moment connections are often utilized in the design of special steelmoment resistant frames (SMRFs). The details of a typical RBS connection are shown in

Figure 1(a), where the flanges of the beam are reduced in width, away from the column face.Complete joint penetration (CJP) groove welds attach the beam flanges to the column. Thebeam web often is welded to the column flange with a CJP groove weld. By design, the RBSconnection develops inelastic deformations primarily in the region where the beam flange

width has been reduced (referred to herein as the RBS), limiting the inelastic strain developedin the beam flange-to-column CJP groove welds. With the reduction of the beam flange width,an RBS connection is more prone to inelastic local buckling of the beam web and flanges inthe RBS. For economical reasons, design engineers in the U.S. prefer to use deep columns

in SMRFs (as large as 914 mm in depth corresponding to a W36 wide flange section) in orderto control seismic drift. Previous tests on RBS connections have been performed primarily oncolumns with depths corresponding to a W12 and W14 wide flange section (Roeder (1)),where the depth was about 305 mm to 356 mm. Some tests using W27 wide flange column

sections (686 mm depth) were conducted by Chi and Uang (2), where the connection was anexterior connection (i.e., only one beam was connected to the column). It was observed inthese tests, that as a result of inelastic beam web and flange local buckling in the RBS, alateral displacement of the beam compression flanges occurs. Shown in Figure 1(b) is the

movement of the compression flanges (the top and bottom flanges of the right and left-handbeams, respectively), where F1and F2represent the beam flange compression forces of the

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two beams. Due to an eccentricity created by the lateral movement of the compression

flanges, a torque is applied to the column. Deep columns tend to have thinner flanges and aweb than a shallower column, resulting in a reduced torsional resistance. Consequently, there

have been concerns that the use of anRBS connection to a deep column in a

SMRF can lead to inferior seismicperformance because of the connectionbeing susceptibility to torsional loadingfrom the beams.

The lack of knowledge of the performance

of RBS beam-to-deep columnconnections under seismic loading led toa study on this topic at Lehigh University(3). The study involved both finite element

analysis and experimental tests. Theeffects of the column depth, a composite

floor slab, panel zone strength, beam webslenderness, and supplemental lateral

bracing at the end of the RBS sectionwere examined. Six full-scale specimenswere subsequently tested involvingdifferent column and beam sizes, a

composite floor slab and supplementallateral bracing. Results and conclusionsfrom the experimental study, along withsome design recommendations arepresented in this paper.

TEST MATRIX

The test matrix for the experimental program is given below in Table 1, where some of thedetails of the six full-scale RBS beam-to-deep column connection specimens aresummarized. All specimens represented an interior RBS connection in a perimeter SMRFwith a composite floor slab, with the exception of SPEC-6 which did not have a composite

floor slab. The parameters investigated in the experimental program included: (1) columnsize; (2) beam size; (3) the floor slab; and (4) supplemental lateral brace at the end of theRBS.

The beam and column section sizes for each specimen were selected on the basis ofintroducing different degrees of torsional effects, predicted by the recommended designprocedure of Chi and Uang (2), while also satisfying the weak beam-strong column criteria inthe ASIC Seismic Provisions (4). The design procedure by Chi and Uang considers the totalnormal stress in the column at 4% story drift due to axial load, flexure load, and torsion. The

predicted total normal stress in the column flange is shown plotted in Figure 2 for variouscolumn sections, including those of the test specimens. Figure 2 indicates that SPEC-2,SPEC-4, and SPEC-5 are predicted to develop column flange yielding. The columns for allspecimens and the beams for SPEC-3 through SPEC-6 were fabricated from A992 steel. The

beams for SPEC-1 and SPEC-2 were fabricated from A572 Gr. 50 steel. Both A992 and A572Gr. 50 have a nominal yield strength of 345 MPa.

Figure 1. (a) RBS connection details, and(b) RBS local buckling and lateral beam

flange movement.

Doubler Plate

R

Erection bolt

F2

F1

e2

e1

(a)

(b)

F2

F1

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Table 1. Test matrix.

SPECColumn

size

Beam

Size

Doubler

Plate

Floor

Slab

Supp.

Lat.

Brace @

RBS

Yield Stress

Flange/Web

(MPa)

Tensile Stress

Flange/Web

(MPa)

Beam Col Beam Col

1 W36x230 W36x1506x800x

1067Yes No 343/378 356/393 478/492 496/514

2 W27x194 W36x15013x610x

1067Yes No 343/378 372/392 478/492 520/502

3 W27x194 W36x15013x610x

1067Yes Yes 365/396 356/403 508/506 497/521

4 W36x150 W36x15010x8160x

1067Yes No 365/396 365/396 508/506 508/506

5 W27x146 W30x10810x610x

914Yes No 344/353 363/399 471/469 499/513

6 W24x131 W30x10813x533x

914No Yes 344/353 334/359 471/469 499/493

CONNECTION AND COMPOSITE FLOOR SLAB DETAILS

The elevation of a typical connection detail is shown in Figure 3. Each specimen wasdesigned in accordance with the criteria recommended by Engelhardt (5) for RBSconnections, where the design moment in the beam at the column face is limited to Mpn,

where Mpn is the nominal plastic capacity of the beam. For the six specimens in the testmatrix, the average value of the beam design moment at the column face was equal to0.973Mpn. The reduction in flange width at the center of the RBS for each specimen was 50%

of the original flange width, which complied with the design criteria of Engelhardt. The RBSwas flame cut, with the burned surface ground to a surface roughness of 500 micro-inches,as recommended by FEMA 353 (6). Each specimen had continuity plates the same thicknessas the beam flanges and designed for a balanced panel zone condition. Complete details are

given in Ricles et al. (3). The weld procedure specifications used in the fabrication of theconnections were prequalified in accordance with AWS D1.1/D1.1M:2002 (7). All welds weredone using the flux core arc welding procedure, and conformed to the AWS 5.20-95Specification (8). The beam flange-to-column flange CJP groove field welds and beam web-

to-column CJP groove field welds utilized E70T-6 and E71T-8 electrodes, respectively. All

shop welds (e.g., shear tab to the column, doubler and continuity plates) were performedusing E70T-1. The run off tabs for the beam flanges were removed following the placement of

Figure 3. Specimen typical connectiondetails (Note: 1 inch = 25.4 mm).

Continuity plate

Doubler plate

E70T-1(TYP)

E71T-8

27"

Shear tab

9"

E70T-1

10"

RBSFlange Cut

Beam

E70T-1

E71T-8

E71T-8

E70T-1

E71T-8

erection bolts

6"6"6"

Column

6"

E70T-1

CJP(TC-U4a-GF)E70T-6

R

CJP(TC-U4a-GF)E70T-6

No Run-off Tabs

CJP

Shear Plate

Figure 2. Column total stress per Chi and Uang(2) versus column section weight.

0

100

200

300

400

500

600

700

0 200 400 600 800 1000 1200 1400Column Section Weight (kg/m)

TotalStress(MPa)

SPEC-4

W36x150

SPEC-2

W27x194

SPEC-1

W36x230

SPEC-3 W27x194(supplemental bracing)

SPEC-6 W24x131(supplemental bracing)

SPEC-5

W27x146

SPEC-5 and 6 had

W30x108 beams; all

others had W36x150

beams


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the CJP groove welds, and the weld at the edges of the beam flanges ground to a smooth

transition. The backing bar of the top flange weld was left in place and a reinforcement filletweld was provided between the bottom surface of the backing bar and the column flangeusing the E71T-8 electrode. The beam bottom flange backing bar was removed using the air-arc process, back gouged, and reinforced with a fillet weld using an E71T-8 electrode. No run

off tabs were used for the vertical beam web CJP groove welds. All CJP groove welds wereinspected using the ultrasonic test procedure in order to evaluate whether they complied withthe criteria in AWS D1.1 (7) for weld quality.

The specimen composite floor slab had a total thickness of 133 mm, and consisted of 27.6MPa nominal compressive strength concrete cast on a 20-gage zinc coated metal deck. AW4x4 welded wire mesh with wire 152 mm on center was placed in the floor slab prior topouring the concrete. The width of the floor slab was 1220 mm to one side, with a 305 mmoverhang on the other side to simulate the conditions for a perimeter SMRF. The ribs of the

decking ran parallel to the main beam (i.e., the beams with the RBS connections) of each testspecimen. To develop the composite action, 19 mm diameter shear studs were placedoutside the RBS region at 305 mm spacing along the beams to attach the deck to the main

beams as well as transverse W14x22 floor beams. These transverse beams were placed at aspacing of 3048 mm to provide lateral bracing to the main beams and column, where thedistance of 3048 mm satisfied the AISC Seismic Provisions (4).

SPEC-6, which had no composite floor slab, had a supplemental lateral brace at the end of

the RBS in addition to the other lateral bracing noted above for the beams. The lateralbracing was attached to a W36x150 section that was placed parallel to the beams of the testspecimen to simulate a parallel beam in the prototype building. This parallel beam in the testsetup was allowed to move horizontally with the test specimen, but restrained from out-of-

plane movement. The corresponding stiffness of the lateral bracing setup satisfied the AISCLRFD Specification (9). SPEC-3 also had supplemental lateral braces, but these were

anchored in the floor slab.

TEST SETUP, LOADING PROTOCOL, INSTRUMENTATION

The test setup is shown in Figure 4 (a), with the lateral bracing detail given in Figure 4(b) forthe main beams. The ends of the members in the test setup had pin-connected boundary

conditions, using cylindrical bearings to simulate inflection points at the beam midspan andcolumn midheight in the prototype frame. The ends of each beam away from the column weresupported by instrumented rigid links, which simulated a roller boundary condition andenabled horizontal movement of the end of each beam. The lateral bracing detail shown in

Figure 4(b) was used to prevent out-of-plane movement of the beams and column (the

diagonal double angles were not used at the column), and designed for strength and stiffnessin accordance with the AISC LRFD Specification (9). The top of the column was bracedagainst torsion, while at the base of the column a clevis was used to create the pin boundary

condition. The beams were also braced at the rigid links in order to stabilize the test setup.The torsional bracing provided at both ends of the column in the test setup was evaluatedusing a nonlinear finite element model (3) to examine whether the stiffness would berepresentative of the torsional restraint at the column inflection points in the prototypestructure. It was found to be satisfactory and not influence the test results by over-restraining

the ends of the column from twisting.

The specimens were tested by imposing a cyclic story drift history based on the loading

sequence defined in Appendix S of the AISC Seismic Provisions (4). The loading protocolconsisted of initial elastic cycles of story drift, followed by cycles of increased amplitude tocause inelastic response. A test was terminated when either a fracture occurred, resulting in


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a significant loss of specimen capacity, or after reaching a story drift of 6%. Each specimen

was instrumented to enable measurement of the applied loads, reactions at the rigid links,specimen story drift; strains in the beam, column, panel zone, and continuity plates; inaddition to panel zone deformation, plastic beam rotation, twisting of the column, and lateraldisplacement of the beam at the center of the RBS.

TEST RESULTS

A summary of test results for each specimen is given in Table 2, where Rv/Vpz, max, Mf/Mpn,

K,col, , flg, and bfare equal to the ratio of panel zone shear capacity-to-panel zone shearforce corresponding to the plastic flexural moment developing in the RBS, specimen drift fromthe last cycle prior to any fracture or strength deterioration to below 80% of the specimen

nominal capacity, ratio of maximum measured beam moment developed at the column face-to-nominal beam flexural capacity, column elastic torsional stiffness, specimen column twistat 4% story drift, lateral displacement of the beam bottom flange at the RBS at 4% story drift,and beam flange width, respectively. Typical observed behavior during the testing of a

specimen consisted of yielding in the RBS and the panel zone, followed by cyclic local weband flange buckling in the RBS. Following the development of local bucking in the RBS,lateral movement of the bottom beam flange began to occur in the RBS of specimens with acomposite floor slab at 2% to 3% story drift. The combined effect of cyclic local buckling andlateral flange displacement resulted in a gradual deterioration in specimen capacity to occur

during subsequent cycles where the story drift amplitude was increased. This is evident in thelateral load-story drift hysteretic response of SPEC-4 shown in Figure 5. The lateraldisplacement of the bottom beam flange occurred when it was in compression, and causedsome column twist to develop. Figure 7(a) and (b) shows photographs of SPEC-4 at 4% and

6% story drift, where the yielding in the members and panel zone in the connection regionand lateral beam flange movement in the RBS are visible. The maximum column twist amongthe specimens with a floor slab at 4% story drift was 0.037 rads. (SPEC-4). 4% story drift isthe drift at which connections are judged for qualification for seismic use by the AISC Seismic

Provisions (4). SPEC-4, like the other specimens, developed a flange fracture in the RBSwhere extensive local flange buckling had occurred (see Figure 7(c)). This occurred at a storydrift of 6%, and was caused by local buckling in the beam flange that led to large cyclic

strains, resulting in a low cycle fatigue failure. SPEC-6, which had a supplemental brace andlateral bracing attached to the beam that is parallel to the test beam, had minimaldeterioration in capacity as well as column twist (0.004 rads. at 4% story drift), see Figure 6.

Figure 4. (a) Test setup and (b) beam lateral bracing detail for specimens with acomposite floor slab (Note: 1 inch = 25.4 mm).

6'-6"

29'-6"14'-9"

Setup Lateral Bracing

Beam Web Stiffener with

Diagonal Brace to Floor

Beam (North Side Only)

Load Cell

14'-9"

LoadCell

1312"

Beam (East)Beam (West)

6'-6"

1312"

10'10'

(North Side Only) (North Side Only)Floor BeamFloor Slab 1312"

No Diagonal Bracing

Floor Beam

21"

ColumnActuator

Load Cell

LCSYM

Floor Slab

W14x22

Shear Stud

48"

A325

3 4" diam.(TYP)

W36x150

514"

12"

Double

Angle

2 L2x2x516

(b)(a)


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The reduced amount of deterioration in the capacity of SPEC-6 was due to the specimenhaving a weaker panel zone than the other specimens. As noted in Table 2, for SPEC-6 the

ratio of Rv/Vpz is equal to 1.03. Rv is based on the ASIC Seismic Provisions (4). All otherspecimens have a value of 1.14 or greater for the ratio of R v/Vpz. Consequently, these otherspecimens developed a larger amount of yielding and local buckling in the RBS than SPEC-6,

leading to local buckling and deterioration in specimen capacity.

In Table 2, all specimens are shown to have a value for max that exceeds 0.04 rads., which isthe current criteria in Appendix S of the AISC Seismic Provisions (4) for qualifying aconnection for seismic use. A summary of the ratio of Mf/Mpn in Table 2 indicates that the

maximum beam moment developed at the column face in the specimens exceeded thedesign value of Mpn for which the specimens were designed, with SPEC-5 having themaximum value of 1.2. The increase in the moment M f is attributed to the composite floor

slab increasing the moment capacity in the RBS. SPEC-6 had a valve of Mfequal to 1.0, andhad no composite floor slab.

Figure 5. Lateral load-story drifthysteretic response of SPEC-4.

-2000

-1500

-1000

-500

0

500

1000

1500

2000

-6 -4 -2 0 2 4 6

LateralL

oad(kN)

Story Drift (% rad.)

Figure 6. Lateral load-story drifthysteretic response of SPEC-6.

-2000

-1500

-1000

-500

0

500

1000

1500

2000

-6 -4 -2 0 2 4 6

LateralL

oad(kN)

Story Drift (% rad.)

(a) Yielding andlocal buckling in

connectionregion, 4% drift

fracture

(c) Beam bottom flange fracture

at RBS, 6% story drift

(b) Beam bottom flange lateral movementat RBS, 4% story drift

Figure 7. Photographs of SPEC-4 during testing.


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Table 2. Test results.

SPECpz

v

V

R max

(% rad) pn

f

M

M K,col

(kN-m/rad)

(1)(rad)

flg(1)

(mm)0.2bf(mm)

1 1.26 4.0 1.03 3190 0.016 53 61

2 1.14 4.0 1.13 1404 0.025 34 61

3 1.28 5.0 1.15 1404 0.006 35 -(2)

4 1.24 4.0 1.11 947 0.037 38 61

5 1.21 5.0 1.20 900 0.007 26 53

6 1.03 4.0 1.00 577 0.004 5 -(2)

Note: (1) Corresponding to 4% story drift(2) Chi and Uang criteria (2) for transverse beam flange movement does not apply to

cases with supplemental braces.

An examination of the results for column twist in Table 2 reveals that column twist tends to

increase when the elastic torsional stiffness of the column K,col is reduced. However, for a

smaller beam section size is reduced, although the column torsional stiffness is smaller

(e.g., SPEC-5). This phenomenon is associated with a smaller demand on the column whena smaller beam is used. The column twist is reduced significantly in specimens with asupplemental brace (SPEC-3 and SPEC-6). The reduction in column twisting in SPEC-6 is

also attributed to a weaker panel zone, which reduced the amount of yielding and localbuckling in the RBS, and subsequently less lateral movement in the RBS. An examination ofthe measured specimen beam flange lateral displacement flgin Table 2 shows these results

to be less than the value of 0.2bf, which is the value recommended by Chi and Uang (2) for

determining the design torque Tapplied to the column. Consequently, the use of the value of0.2bffor determining the design torsional loading on the column from the RBS will result in a

larger column design torque. This is evident by comparing the column total normal stress atthe connection based on Chi and Uangs recommendation with the measured specimenresponse (see Table 3). The criterion by Chi and Uang anticipates column flange yieldingoccurring in SPEC-2, 4, and 5; see Figure 2, where the nominal yield stress is 345 MPa. Themeasured column flange longitudinal strains in these specimens indicated no yielding in

SPEC-2 and 5, with some minor yielding occurring in SPEC-4 (a maximum strain of 2 to 4times the yield strain developed). The measured longitudinal strains across the column flangejust below the connection are shown in Figure 8 for SPEC-4 and SPEC-6. These results are

-6000

-4000

-2000

0

2000

4000

6000

-150 -100 -50 0 50 100 150

Strain(microstrain)

Distance across column flange (mm)

SPEC-4SPEC-6

Figure 8. Longitudinal strain profile across column flange, just belowRBS connection; 4% story drift.

=1765


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representative of typical specimen behavior, and show little evidence of a strain gradient

across the flange that would result from the effects of warping normal stresses due to columntorsion.

DESIGN RECOMMENDATIONS

The strains in the beam bottom flange near the column face were examined to evaluate the

stress distribution across the beam flange that leads to a torque Tapplied to the column.Shown below in Figure 9(a) is the distribution of longitudinal stress across the beam bottomflange at 4% story drift. These stresses are based on measured longitudinal strains in thespecimens. These results correspond to a negative beam moment at the column face (i.e.,

when the bottom flange of the beam is in compression). Similar results for longitudinalstress across the beam compression flange were obtained from finite element studies (seeFigure 9(b)). The results in Figure 9 show a trend where the stress distribution across thebeam flange has a reduction in stress, which is due to a moment in the plane of the beam

flange caused by the lateral movement of the beam flange at the RBS. This moment is

equivalent to the torque Tthat is applied by the beam flange to the column. Shown in Figure10(a) is an idealized uniform longitudinal stress distribution prior to lateral movement of thebeam flange in the RBS (at 2% story drift). The idealized longitudinal stress distribution at 4%story drift based on the measured and finite element analysis results is given in Figure 10(b).At 4% drift local buckling and lateral beam flange movement has occurred in the RBS.

Elastic-perfectly stress-strain behavior is assumed in Figure 10, where Fyeis the yield stress.

-500

-400

-300

-200

-100

0

100

-150 -100 -50 0 50 100 150

Longitudinals

tress(MPa)

Distance across beam flange (mm)

-400

-300

-200

-100

0

100

200

300

400

-150 -100 -50 0 50 100 150

Stress(

MPa)

Distance across beam flange (mm)

SPEC-1SPEC-2SPEC-3SPEC-4SPEC-5

Figure 9. Longitudinal stress distribution across beam flange for (a) all testspecimens, and (b) finite element analysis of SPEC-2.

(a)

(b)

(a) (b)

Figure 10. Idealized longitudinal stress distribution across beam bottom flange at(a) 2% story drift and (b) 4% story drift.

Fye Fye


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For the longitudinal stress distribution shown in Figure 10(b), Tcan be shown (3) to be equalto

f

2

fye tbF150

11T= (1)

where Fye, bf, and tfare equal to the expected beam flange yield stress (1.1Fy), the beamflange width, and beam flange thickness, respectively.

A design procedure was thus developed in order to determine the total design longitudinal

stress ftotal in the column flange that is attached to an RBS connection. The procedureinvolves determining the elastic warping normal stresses fwthat develop in the column flange

due to the torque T (10) and superimposing them with the column flange normal stresses

due to bending (fb) and axial loading (fa) to obtain the total normal stress ftotal, where

"

w nOEWf = (2)

In Equation (2) E, WnO, and are equal to the Youngs modulus, normalized warpingfunction at the column flange tip (10), and the second derivative of the angle of twist in the

column (10), respectively, where is a function of the torque T.

The total stress ftotalis compared to the criteria in the AISC LRFD Specification (9), Equation(H2-1), where

yFf =total (3)

in which and Fy are the resistance factor (0.9) and nominal yield stress of the column

flange, respectively. The above design procedure is similar to that developed by Chi and

Uang (2), except for the method in which the torque is determined.

Table 3. Comparison of column normal flange compression stresses with design procedure.

Warping

stress fw

(MPa)

Total normal

stress ftotal(MPa)

Experimental

results, total

stress & strain,

4% story driftSPEC Column Beam

Axial

load

stress

fa

(MPa)

Bending

stress

fb(MPa) Chi

and

Uang

Pro-

posed

Chi

and

Uang

Pro-

posed

Strain

()

Stress

(MPa)

1 W36x230 0 190 128 66 318 256 1277 255

2 W27x194 0 299 182 101 481 400 2151 372(1)

3 W27x194 0 332 0 0 332 332 1797 356(1)

4 W36x150

W36x150

0 337 321 163 658 500 3296 365(1)

5 W27x146 0 252 180 95 432 347 1598 319

6 W24x131W30x108

0 347 0 0 347 347 2525 334(1)

Note: (1) Yield stress of the column flange.

The total normal column flange stress based on the above procedure is compared in Table 3to the measured stress of the test specimens, as well as the stress predicted using the

procedure by Chi and Uang (2). The comparisons in Table 3 indicate that a more accurateprediction of the total normal stress in the column flange is made using the above procedurecompared to the procedure developed by Chi and Uang (2). The difference between the two


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methods is the normal warping stress fwpredicted by the above procedure is based on a

more accurate value of the torque T applied to the column. For specimens with asupplemental lateral brace it was assumed that the restraint of the supplemental braceresulted in no torque applied to the column (i.e., the normal warping stress fw is equal to

zero). This results in a lower predicted stress than the measured response.

SUMMARY AND CONCLUSIONS

An experimental program was conducted in order to evaluate the seismic performance ofRBS connections to deep wide flange columns. The study involved testing six full-scalespecimens to evaluate the effects of column depth, beam size, composite floor slab, and a

supplemental lateral brace.

Based on the experimental study, the following main conclusions are noted:

1. A composite floor slab can significantly reduce the lateral displacement of the beam

bottom flange in the RBS and the amount of twist developed in the column. The slabappears to be effective in reducing the twist in deeper columns attached to an RBSconnection, and enables the cyclic strength of the beam with an RBS connection tobe better sustained.

2. All of the specimens were able to satisfy the criteria in the AISC Seismic Provisions(4) for qualifying the connection for seismic use.

3. A weaker panel zone in a deep column RBS connection will not develop as muchcolumn twist and strength degradation as a connection with a stronger panel zone.

However, a weaker panel zone can significantly increase the potential for ductilefracture of the connection (3). It is recommended that connections be designed with abalanced panel zone strength condition.

4. A supplemental brace at the end of the RBS significantly reduced the transversemovement of the beam flanges in the RBS and column twist that leads to cyclicdegradation in specimen capacity.

5. Basing the column torque on a transverse movement of the beam flange in the RBSof 0.2bffor calculating column flange warping stresses appears to be conservative. A

new procedure for estimating the torsional load applied to the column due to the localand lateral buckling in the RBS shows improvement in predicting the correct columnflange normal stress.

ACKNOWLEDGEMENTS

The research reported herein was supported by a grant from the American Institute of SteelConstruction (Mr. Tom Schlafly program manager) and from the Pennsylvania Department of

Community and Economic Development through the Pennsylvania Infrastructure TechnologyAlliance (PITA) program. The following companies donated materials for the experimentaltesting conducted in this research project: Arcelor International America of New York, NY(steel sections); Nucor Vulcraft Group of Chemung, NY (metal decking); and the Lincoln

Electric Company of Cleveland, OH (welding wire). The support provided by the fundingagencies and companies is greatly appreciated.


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REFERENCES

(1) Roeder, C. W. (2000). Connection Performance State of Art Report, Report No.

FEMA-355D, FEMA, Washington, D.C.

(2) Chi, B. and Uang, C.-M. (2002). Cyclic Response and Design Recommendations of

Reduced Beam Section Moment Connections with Deep Columns, Journal ofStructural Engineering, ASCE, 128(4): 464-473.

(3) Ricles, J., Zhang, X., Lu, L.W., and J. Fisher, (2004). Development of Seismic

Guidelines for Deep-Column Steel Moment Connections, ATLSS Report No. 04-13,ATLSS Engineering Research Center, Lehigh University, Bethlehem, PA.

(4) Seismic Provisions for Structural Steel Buildings, (2002). American Institute of SteelConstruction, Chicago, Illinois.

(5) Engelhardt, M. D. (1999). The 1999 T. R. Higgins Lecture: Design of Reduced Beam

Section Moment Connections, Proceedings: 1999 North American Steel ConstructionConference, American Institute of Steel Construction, Toronto, Canada, pp. 1-1 to 1-29.

(6) Recommended Specifications and Quality Assurance Guidelines for Steel Moment-

Frame Construction for Seismic Applications, (2000). Report No. FEMA 353, FederalEmergency Management Agency, Washington D. C.

(7) Structural Welding Code Steel, (2002).AWS D1.1/D1.1M:2002, American WeldingSociety, Miami, Florida.

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