Incercari in Situ - Utilitate Si Interpretare

7/28/2019 Incercari in Situ - Utilitate Si Interpretare

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50 INTERPRETATION OF IN SITU SOIL TESTS

clay being directly measured in situ, or moni-tored indirectly by the decay of excess porewa-ter pressures.

The fourth purpose, the monitoring of perfor-mance of geotechnical works, may be a standardprocedure such as the continuous observation ofmovement, of porewater pressures and of quan-tities of seepage in an earth dam, or in cir-cumstances where there are special problems oruncertainties. A striking example of the latterwas the monitoring and subsequent backanalysis of the New Palace underground carpark at the Houses of Parliament, London, re-ported by Burland & Hancock (1977).

Most calculations carried out in the past bypractising civil engineers for the design of foun-dations and earthworks have been restrictedeither to limit analysis for stability calculationsor to predictions of settlement. Classical limitanalysis is independent both of the deformationcharacteristics of the ground and of the level ofthe in situ lateral stress of the undisturbedground; it depends solely on the groundwaterconditions and on the properties of strength andunit weight of the soil or rock. In contrastclassical methods of settlement prediction de-pend only on deformation properties.

In most instances the relevant properties havebeen evaluated from laboratory tests on sup-posedly undisturbed samples, and then beenused in a simple analysis to lead to designs whichhave proved to be entirely satisfactory. Why,then, is it necessary to make in situ measure-ments of soil and rock properties?

The main reason for this need is that, as ourknowledge of the behaviour of real soils in-creases, so our appreciation of the inadequacy ofconventional laboratory testing grows. Themarked consequences of the inevitable distur-bance that is caused in any soil specimen, how-ever carefully it has been sampled, transportedand reconsolidated in the laboratory, are all tooevident. The work at the Building ResearchStation has shown, for example, that the actualdeformation moduli of the ground may be sev-eral times greater than those measured in goodquality tests in the laboratory on good qualitysamples, as shown e.g. by Marsland (1973).Consequently predictions of the deformation ofthe ground around a foundation or excavationbased on laboratory data may be grossly overes-timated, and the resulting design may be un-necessarily conservative and expensive.

A separate but important advantage of in situtesting is that the soil in question will be testedat the appropriate level of effective stress, pre-suming that disturbance of the ground due toinsertion of the instrument has been kept to aminimum.

Apart from good technical reasons for con-ducting in situ tests, there may be situationswhere the total cost of site investigation andtesting makes them economically attractive, orwhere they must form the major part of theinvestigation such as in the exploration ofoffshore sites for oil production platforms.

In parallel with the major developments thathave occurred in the last 25 years or more inexperimental techniques, in instrumentation andin the understanding of soil behaviour have beenthe profound changes in analytical methodsmade possible by the electronic computer. Newmethods of numerical analysis not only allowcomplete solutions to be obtained to complexboundary value problems but also allow the useof non-linear, non-homogeneous, anisotropic-and hence more realistic-models of soil or rockbehaviour.

In the past few years there has been a markedgrowth in the use of in situ tests and in thevariety of instruments that have reached a suffi-ciently developed stage that they can be usedwith confidence. It is not possible within thelimits of this Paper to attempt a comprehensivereview of these instruments or of the currentstate of in situ testing. The purpose of the Paperis to discuss the interpretation and use of theresults of in situ tests, and to highlight some ofthe considerable difficulties and uncertainties as-sociated with them. Most of the discussion isconcentrated on(a) in situ tests in clay (most of the principles

involved will apply to other soils and rocksto a greater or lesser degree)

(b) results of self-boring pressuremeter testsand piezocone tests.

The reasons for this choice are given later.

RELATIONSHIPS BETWEEN SOILPROPERTIESThe interpretation of data obtained from insitu tests is difficult, and for most tests it is bothincomplete and imprecise. A number of separatefactors contributes to this unsatisfactory situa-tion. The factors fall into two distinct categories:those due to the behaviour of the soil and thosedue to the type of test being performed.Soil behaviour is complex and depends on thecomplete geological history of the deposit asrepresented by the size, shape, mineral composi-tion and packing of the particles, the stresshistory that has been experienced, the pore fluidand other factors. The response of the soil to aparticular test will depend on the changes ineffective stress that it undergoes, and, further,this response will be inadequately represented


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WROTH 451by a few simplistic properties such as undrainedshear strength, shear modulus, coefficient ofconsolidation etc. The properties themselvesmay vary locally to a significant degree bothlaterally and vertically within the ground, owingto the microfabric of the material and the quirksof its history.

Any in situ test, when considered as a bound-ary value problem, is beset with difficulties. Theboundaries of the problem are unknown anduncontrolled, so that there are insufficient datafor a complete solution and for an unequivocalinterpretation of the results. The fields of stressincrement and strain induced around the instru-ment by the operation of the test vary signific-antly with distance from the instrument; thisvariation is not unique for the type of test but isitself dependent on the stress-strain propertiesof the soil being tested.

In all but fully drained situations, the non-homogeneous fields of stress cause locally highhydraulic gradients so that some degree of par-tial consolidation will occur in what is supposedto be-or is interpreted as-an undrained test.This partial consolidation introduces an impor-tant rate effect, in addition to that attributableto the viscous nature of soil behaviour.

A further complication arises in that in all insitu tests (except the pressuremeter test) theprincipal axes of stress rotate within the soil,whereas they do not in the triaxial test, which isused as a standard form of comparison.

In addition, in all experimental work there arelimits to the accuracy and reliability of the in-strument, a situation which is worse in the fieldthan in the laboratory.

Consequently any interpretation of an in situtest is open to question. To make the most ofthe interpreted results it is vital to correlatethem with the results of all other data, whetherfrom the field or the laboratory, and to draw onall available experience.

The choice of properties that should be usedin any attempted correlation is crucial. Anysuccessful relationship that can be used withconfidence outside the immediate context inwhich it was established should ideally be(a) based on a physical appreciation of why the

properties can be expected to be related(b) set against a background of theory, however

idealized this may be(c) expressed in terms of dimensionless vari-

ables so that advantage can be taken of thescaling laws of continuum mechanics.

An illustration of these points is provided byconsidering correlations of the undrained shearstrength s,, of a clay. All soils are basicallyfrictional materials with the strength being pro-

vided by the frictional resistance between soilparticles governed by the effective stress towhich they are subjected. Starting ab in i t io , thefirst relationship to be explored would be

$-f(4)where pfl is the mean principal effective stress atfailure and 4 i s the angle of shearing resistance.In any real situation the value of pr will not beknown, and it has to be replaced by some otherstress variable. If the initial conditions areselected, and the mean principal effective stressp,, is used, then its relationship with pr) dependson the excess pore pressures generated duringshearing to failure, which in turn depends on theoverconsolidation ratio OCR of the clay. Hencea second approach would be to consider therelationship

5 = (4, OCR)PC,However, this relationship will in practice besubject to much uncertainty because the in situmean principal effective stress p,, is unlikely tobe known or to have been estimated with anyaccuracy. The single stress variable that can beestimated with most reliability is the in situvertical effective stress oVo. Since this is relatedto p,) as a function of OCR, its use in lieu of pO will not increase the number of variables in therelationship. Consequently a good engineeringcompromise is to adopt the expression

2 = f (4,03-Q (3)and to define s,/u,~, as the undrai ned strengt hratio.

Historically in soil mechanics, much use hasbeen made for normally consolidated clays ofthe relationship suggested by Skempton (1957)

S- = 0.1 1 + @0037PIovO (4)where PI i s the plasticity index of the clay.Within the context of these arguments, is this asound relationship, and does it suggest a newvariable PI that should be taken into account?At first sight, it is not evident that the undrainedstrength ratio should be related directly to theplasticity index. However, the value of 4 can beexpected to depend on the shape, size, packingand mineral composition of the clay particles, aswill the plasticity index, so the two propertiesare related in some complex manner. Thus phys-ical reasoning supports Skemptons relationshipbut suggests that it would be a weaker one than


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452 INTERPRETATION OF IN SITU SOIL TESTSthat of equation (3) in which 4 is preferred toPI.CONDITIONS AT FAILURE

The majority of in situ tests induce local fail-ure in the soil, and the most commonly deducedproperty is the undrained shear strength. Muchof this Paper is therefore taken up with a de-tailed and critical look at undrained shearstrength.

The symbol used in this Paper for undrainedshear strength is s,, in accordance with commonpractice in the USA, rather than the symbol c,as recommended by the British Standards In-stitution (1975). This is a deliberate choice, be-cause the former relates to strength whereasthe latter relates to cohesion; it is argued inthis Paper that strength must be interpreted interms of effective stresses and friction angle, andnot total stresses and cohesion.

The basic definition of undrained shearstrength is

S = +(a, - o.3) (5)i.e. half the difference between the major andminor principal stresses, or the radius of thelargest Mohr circle. This is an unsatisfactorydefinition as it neither takes account of theintermediate principal stresses o2 nor distin-guishes between the different types of test whichare well known to give different results for iden-tical soil specimens.It is essential to distinguish between differenttest results by an inelegant plethora of suffices asfollows:SW,S teSpS.3SlmSdSS%f&llScP

triaxial compression testtriaxial extension testplane strain active testplane strain passive testdirect simple shear test

laboratory

field vane testpressuremeter test fieldcone penetrometer test

The major question to be faced is how thesedifferent measurements of strength are con-nected. An attempt is made to link them bymeans of the friction angle 4.

The basic concept of an angle of shearingresistance or of internal friction comes from theclassical experimental work of Coulomb inwhich (plane strain) tests were conducted in ashear box. The resulting definition of the frictionangle in terms of principal effective stresses is

Fig. 1. Sections of failure surfaces in principal stressspace

Unfortunately, as for undrained shear strength,this neither allows for the influence of uz nordoes it distinguish between different types oftest. Because it is absolutely essential to under-stand soil behaviour in terms of effective stres-ses, and not total stresses, the use of the symbol4 should be abandoned and the use of theprime in the symbol 4 can be dropped, both forconvenience and for emphasis.

Furthermore, to make a proper comparisonbetween failure conditions in different tests, thelink will be established initially between resultsof tests on nomally consolidated clays, so thatthe friction angle relates not to peak strengthbut to conditions at the end of a test when thereis no further change in volume or effective stres-ses, i.e. at the ctitical state. Hence 4 replaces6, or 4-l.Suffices are required to distinguish results, asfollows:

t 1:triaxial compression testtriaxial extension test4,. plane strain tests.

On the basis of limited experimental evidence itis assumed that the critical state friction angle isthe same for all plane strain tests includingdirect simple shear tests.To take proper account of the effect of theintermediate principal effective stress u2 it isnecessary to adopt a generalized failure criterionexpressed in terms of all three principal effectivestresses. Fig. 1 is a section of principal stressspace made by a rr plane or octahedral plane,which is perpendicular to the space diagonal.The figure shows the sections of three possiblefailure surfaces, each of which emanates fromthe origin in the form of a cone. The inner


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WROTH 453irregular hexagon is the classical extendedMohr-Coulomb failure envelope given simplyby the requirement that

4 = constant (7)The outer, broken curve is a section of thefailure surface proposed by Lade (1972). Since

this criterion is expressed in terms of all threeprincipal stresses it is best to make use of thestress invariants

and to write Lades criterion as113/13 constant (9)An alternative failure criterion is that prop-

osed by Matsuoka (1974) which has the forml,I,/I, = constant (10)

and which is represented by the full, inner curve.Both sections of the Lade and Matsuoka fail-

ure surfaces have been drawn so that theycoincide with the Mohr-Coulomb criterion fortriaxial compression tests (i.e. they pass throughthe vertices of the hexagon which lie on thepositive stress axes). The Matsuoka curve alsopasses through the other three vertices of thehexagon, whereas the Lade curve does not. Thetwo criteria are very similar, and the curves havethe subtle property that the shape varies withthe friction angle; as 4 decreases the shapebecomes more circular, and as 4 increases theshape becomes more triangular.

Of the three criteria, Matsuokas is chosen forthe current analysis for three reasons.(a) It was initially developed from theory and

not from curve fitting of experimental data.The theory, which is complex, is based onthe concepts of spatially mobilized planeswithin a soil specimen on which slip is as-sumed to occur (Matsuoka & Nakai, 1977).

(b) It appears to fit experimental data best.*However, the data from complicatedlaboratory apparatus in which the principalstresses can be independently controlled arenotoriously suspect.(c) It is expressed in terms of all three stressinvariants.

*Reference can be made to Matsuoka & Nakai(1982), where data are presented for sand from theresults of their tests, of Sutherland & Mesdary (1969)and of Ramam urthy & Rawat (1973), and for clayfrom Shibata & Karube (1965).

A priori, there is no reason to omit the secondinvariant IZ, as Lades criterion does.

Bishop (1966) introduced the parameter'--U3'b=+-

Cl u3(11)

as a convenient way of expressing the relativevalue of the intermediate principal effectivestress. The value of b varies between zero fortriaxial compression and unity for triaxial exten-sion. Its value for plane strain conditions hasbeen much debated, but it has often been takenas 0.5 on the basis of the theory of perfectplasticity.

For the particular case of triaxial compressionfor which ~2) = Us (and b = 0) Matsuokas criter-ion can be expressed as

I,I, (Ul + 2u3)(u3* + 2u11u3)-=I, o1o3 ,2= (u1/u3(+ 2)(1+ 2u,/u3)

UI/U3 (12)which can be rewritten in terms of &, as fol-lows:

I,I, (3-sin &)(3+sin &)r,= (1 -sin &)( 1 + sin &,)

=9+8tan& (13)For triaxial extension, an exactly similar deriva-tion leads to

I,& (3+sm &)(3-sm &)I,= (1 + sin 4,)( 1- sin 4,,)

=9+8tan2& (14)which confirms that for this criterion dtC= 4,e.

It is required to relate, if possible, 4,, with 4,,by some simple relationship. Satake (1982) hasshown that if an associated flow rule is appliedto the Matsuoka failure criterion (treated as ayield surface) then for plane strain conditions4pS is the maximum value that 4 can have (forall values of b). This assumption of an associatedflow rule is not valid for peak conditions but issuggested as acceptable for critical state condi-tions.

By finding the maximum value of the ratiou1/u3 (i.e. the maximum value of 4) for a fixedvalue of I1 (i.e. for one octahedral plane) it canbe shown that 4, (~4,~) is given by

set 4p,. +set 4,, =2 set 4t, (1%and further that the value of b for plane strain

bps sin 4, +cOs 4,,- 12 sin 4,. (16)


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454 INTERPRE TATION OF IN SITU SOIL TESTS

andAb=O

Fig. 2. Plane strain conditions derived from Mat-suokas criterionor alternatively

u2( )l+(Tg PS= f cos 4,. (17)These findings are illustrated in Fig. 2 where Pindicates the point on the failure surface corres-

ponding to plane strain conditions. Note thatCD is a line of constant 4 (the MohrCoulombsurface), but that the orientation of such a linechanges slightly with the value of 4, so that thetangent at P associated with 4,, will not beparallel to CD. Taking for example a value ofsin 4,= 0.6 (4,, = 36.87) then 4,, = 32.5 andb,, = l/3.

The relationships of equations (15) and (16)are plotted in Fig. 3. The resulting curves can beapproximated for engineering purposes by thelinear relationships

84,: = 94,: (18)Qp,=: 9@,c

(19)Equation (18) will be used for relating the re-sults of plane strain tests with triaxial compres-sion tests for a given soil.UNDRAINED TRIAXIAL COMPRESSIONTESTS

The triaxial compression test has become thestandard method of obtaining stress-strain andstrength properties of soils in the laboratory aspart of a conventional site investigation. It isagainst such a background that the results of insitu tests will be judged in general. Conse-quently it is valuable to establish a theoreticalunderstanding of the triaxial test, and it is be-lieved that this is best done by means of theframework provided by critical state soilmechanics (CSSM).

The original concepts of this approach to soilbehaviour were based on the idea of a criticalvoid ratio conceived by Casagrande (1936), theearly triaxial tests on sand by Taylor (1948) andthen extended by a detailed analysis of triaxialtests on isotropically consolidated specimens ofreconstituted clay by Roscoe, Schofield & Wroth(1958).

In Fig. 4 is presented a comparison betweenthe classical representation of one-dimensionalconsolidation of Terzaghi and the modern ap-proach for isotropic consolidation of CSSM. Theformer has voids ratio e plotted against log,, (r,,whereas the latter has specific volume V plottedagainst log, p, with the consequence that thecompression and swelling indices are defineddifferently.

In one-dimensional swelling the principalstresses do not decrease proportionally, so that

2bpsz1 ~ @ps' / l OO

b = 0.5

:-I//; , ;., , , , y D10 20 30 40 10 20 30 40 50 %s

@ o,C 0 10 20 30 40 @tcOFig. 3. Relationships for plane strain conditions derived from Matsuokas failmecriterion


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WROTH 455

One dlmenslonal. uv Isotropic. p = (0,+ a*+ 0,)/3CC cs A.KOCR=0 /c Overconsolldalsonvmax ratio R = P,,,IPI

uf2 Equivalent pressure(Hvorsiev) P,

Terraghi CSSM

Fig. 4. Defhitions of consolidation in one-dimensional and isotropicconditions

care is required with the definition of the over-consolidation ratio which will not be the same inthe two plots. For isotropic conditions the over-consolidation ratio is defined as the ratio of themaximum past mean effective stress pm_ to thecurrent value p, and it is given the symbol R inaccordance with Atkinson & Bransby (1978).

The equivalent pressure of a specimen, intro-duced by Hvorslev (1937), is defined as thepressure on the normal consolidation line, suchas at point E, at the same voids ratio as that ofthe specimen at state D. This proves to be anelegant and convenient way of converting fromthe voids ratio (or water content) of a clayspecimen into a pressure variable for compari-sons in dimensionless form.

The idealized results of undrained triaxialcompression tests are represented in Fig. 5 interms of the state variables p, q and V. Aspecimen initially normally consolidated at pointC undergoes the effective stress path CDwhereas an initially overconsolidated specimenat point R experiences the path RS; it is as-sumed that both specimens reach undrained fail-ure on the critical state line at D and S respec-

tively. The critical state line is assumed to beparallel to the isotropic normal consolidationline ABC in the semilogarithmic plot, and itsrelative position to be given by the spacing ratior defined as the ratio of the pressures at C and Xwhich lie on the same swelling line CXR. Forthe original Cam clay model r = 2.718 (= e, thebase of natural logarithms), whereas for themodified Cam clay model r = 2.In Appendix 1 it is shown that

p./p, = (R/r) (20)where A = (A - K)/A. This parameter was intro-duced by Schofield & Wroth (1968) because itplays an important role in realistic elasto-plasticmodels of soil behaviour which incorporatestrain-hardening plasticity. It may be termed theplastic volumetric strain ratio, being the ratio ofthe plastic component to the total component ofthe volumetric strain increment in normal con-solidation. This parameter A consistently ap-pears as an exponent in the subsequent analyses.The undrained shear strength in compressionis half the deviator stress at failure so that

s,t, = $4. = ;Mp, (21)


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L

Overconsolidatlon ratio: R = p,Ip,Spacing ratio of ICL and CSL: r = p,Ip, (Z 2)Undrained strength ratlo. su,,/p= MI2(RIr)Normallzed strength ram (su,,lp)l(s,, uc lp),c = RA

Fig. 5. lkoretical expressions for undrained strength in tdxialcompression tests

Substituting for p. from equation (20) givessutc M R A_=_ -0l 2 r (22)

where for triaxial compressionM = 6sin ,

3 -sin & (23)This expression is valid for specimens that havebeen isotropically consolidated, so that at thestart of the compression test uVO pO. Henceequation (22) can be reinterpreted as an expres-sion for the undrained strength ratio

sutc M R A_=- -U 0 rFor any given soil M , r and A will be constantsso that in theory the undrained strength ratio isproportional to the overconsolidation ratioraised to the power A. Evidence in support ofthis finding is presented later.Experimental evidence supporting these con-cepts is provided by the results of undrainedtriaxial compression tests on isotropically con-

solidated specimens of reconstituted kaolin byLoudon (1967). The effective stress paths for aset of specimens are presented in Fig. 6 wherethe stresses have been made dimensionless bydividing by the relevant value of the equivalentpressure pe in each case. In this plot the criticalstate line is reduced to a single unique crit icalstate point indicated by C.In the development of the understanding ofsoil behaviour, the major centres of experimen-tal research in soil mechanics have wisely con-centrated their efforts on testing a limited rangeof soils in a comprehensive manner. Examplesthat come readily to mind are kaolin, Wealdclay, London clay, Boston blue clay and Dram-men clay. This philosophy of research has al-lowed a reliable and detailed picture to be builtup of the behaviour of these clays but has theminor disadvantage that there has not been asurvey of trends of soil properties for a widerange of clays. In particular there are unfortu-nately few reliable data on the range of values ofthe spacing ratio r and the plastic volumetricstrain ratio A. On the basis of the limited evi-dence available it seems that neither of theseparameters varies significantly for a wide range


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457

PPFig. 6. E ffective stress path for undr ainedbiaxial compressiontests on kaolin (afterL mdon, 1967)

of clays, and that for present purposes it isreasonable to adopt the approximate values

r=2A=@8 I (25)

Use of these values means that the undrainedstrength ratio for normally consolidated speci-ments (R = 1) can be rewritten by combiningequations (23) and (24) to give

= 0.5743 3 sin 4,,3 - sin & (26)The variation in this ratio with the friction angle& is shown in Fig. 7 by the upper curve markedITC (denoting Isotropically consolidated speci-mens tested in Triaxial Compression). Note thatfor a clay that has values of r and A that differfrom those assumed (equation (25)) the conse-

0: @lC15" 20" 25" 30" 35"J @20" 25" 30" 35" 40" p*

The concepts of CSSM require that the stateof an anisotropically normally consolidatedspecimen of clay lies on the state boundarysurface at some point such as B in Fig. 6 (refer-ence can be made to Schofield & Wroth (1968)or Atkinson & Bransby (1978)). It is assumedthat failure of this specimen will occur at thecritical state given by point C, with the sameundrained shear strength as an isotropically nor-mally consolidated specimen at the same watercontent (i.e. having the same equivalent pres-sure). However, because the initial stress statesare different, the undrained strength ratios willbe different; the actual value for the anisotropicspecimen will depend on the shape of the stateboundary surface, as well as on the value of KO.

Fi g. 7. Vari ation in the undrainedstrength ratio w ith In Appendix 2, it is assumed that the statethe angle of friction for hiaxial tests on normally boundary surface is formed by the elliptical yieldconsolidated day surface of modified Cam clay. The analysis gives

quence is merely one of scaling the ordinate axisappropriately: the shape of the curve wouldremain unchanged. For later comparisons thevalues of the friction angle in plane strain &,.given by equation (18) have been included onthe axis.Specimens consolidated one dimensionally

In nature, it is usually assumed that soildeposits have become consolidated underone-dimensional conditions. To simulate soil be-haviour in the laboratory, it is becoming increas-ingly the practice to reconsolidate specimensanisotropically to their assumed in situ stresses.The behaviour of such specimens will not be thesame as that of isotropically consolidated speci-mens; this difference must be accounted forwhen the results are used for predicting fieldbehaviour.


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the following cumbersome expression for theundrained strength ratio:

3 -sin &,a = 2(3 - 2 sin +,,)

(27)

For A = 0.8, this expression has been plotted inFig. 7 as the lower curve marked K,TC.

There is a substantial difference between thetwo curves, especially for higher friction angles,so that it is important when undrained strengthratios obtained in triaxial tests are quoted that itis clearly stated whether the specimens wereinitially isotropically or anisotropically consoli-dated. This distinction is important for all realsoils, even if the idealized model of modifiedCam clay is not deemed to be relevant.Overconsolidated specimens

The undrained strength ratio for overconsoli-dated specimens is fully specified by CSSMtheory and has been expressed by equation (24),i.e.

sutc M R _=- -co 0 rThe fact that all the effective stress paths in Fig.6 approach and nearly reach the unique criticalstate point C means that this relationship isclosely followed by the idealized situation ofLoudons tests on isotropically consolidatedspecimens of reconstituted kaolin.

For testing its application more widely it isbetter to normalize the undrained strength ratioin the form

s,&v,(St,IuO)), R (28)

20 1 A 0.8?5bZ 4 Drammen clayX

;t_jl__l__

.x

2

I I4 8 10 20 40OC R

Fig. 8. Variation in the normafized undrained strength Fig. 9. Variation in the normalied undrained strengthratio with overconsolidation ratio for Drammen clay in ratio with the overconsolidation ratio for various clayssimple shear (data from Andresen et al., 1979) in simple shear (data from Ladd & Edgers, 1972)

which has the advantage of being independentof both the frictional coefficient M and thespacing ratio r. This is a powerful relationshipwell supported by various sets of data, each ofwhich refers to one type of test on one particularclay.The most comprehensive sets of tests havebeen carried out at the Massachusetts Instituteof Technology where direct simple shear testshave been conducted on a range of clays in theGeonor apparatus (Ladd & Edgers, 1972). Dur-ing such a test the maximum value of shearstress applied to the specimen, T,,;,,, is observed.The specimens have necessarily been consoli-dated one dimensionally so that the directlyanalogous expression to equation (28) is

(TIz;o;;;;,,c = (OCR) (29)The results of tests carried out on Drammen

clay at the Norwegian Geotechnical Institute byAndresen, Berre, Kleven & Lunne (1979) havebeen replotted in Fig. 8 with both variablesplotted with logarithmic scales. These data havebeen used because the tests are the only onesknown to have included specimens with valuesof OCR as high as 40. Equation (29) suggeststhat the results should lie on a straight line withgradient A, which is indeed the case with A=0.8.

Similar data for seven different clays assem-bled by Ladd & Edgers (1972) were presentedin the state of the art report by Ladd, Foott,Ishihara, Schlosser & Poulos (1977) at the Inter-national Conference on Soil Mechanics andFoundation Engineering at Tokyo. The bounds tothose curves are presented in Fig. 9 in which the

(im,,//cima~!o,l,, = IOCR)6-

with A = 0.8 range ofvalues5- lor 7 clays

4-


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WROTH

Table 1. Values of exponent m for equation (30) for five clays (data from Ladd & Edgers, 1972)LL:% PL: %

LL, liquid limit; PL, plastic limit.normalized undrained strength ratio is plottedwith an arithmetic scale, but the overconsolida-tion ratio has a logarithmic scale.

Ladd et al. (1977) suggest that the data can bewell defined by the expression (using their sym-bols)

part of a comprehensive state of the art reviewof in situ testing. The tests have been listed bythe authors in what they consider to be anascending order of cost and/or complexity.

(c/~)c(ClU),= (OCR)m (30)

The list is by no means exhaustive, omittingfor example(a) push-in pressuremeter (Henderson, Smith &

St John, 1979)with m = 0.8, though a better fit is obtained ifm is decreased from 0.85 to 0.75 with increas-ing OCR. Their finding, which was solely basedon empiricism, is confirmed by the CSSMtheory. Moreover CSSM theory relates the ex-ponent m with well-recognized physical proper-ties of the clay in question by stipulating thatm =A, the plastic volumetric strain ratio.

(b) push-in spade-shaped pressure cell (Tedd &Charles, 1981)

(c) self-boring permeameter (Baguelin,JCzCquel & Le Mehaute, 1974)

(d) self-boring plate test (Mori, 1983)(e) electrical conductivity probe (Arulanandan,

1977).

The original data from the report by Ladd &Edgers (1972) have been replotted in the man-ner of Fig. 8, and a least-squares regression

. analysis has been performed to find the best fitof a straight line to each set of data. As Laddhas pointed out, the data do not lie perfectly ona straight line in this plot, but consistently on ashallow convex curve such that a reduced valueof m is required for large values of overconsoli-dation. Accordingly a distinction has been madebetween excluding all tests with OCR27 andobtaining a value for m = m,, say, and includingall data and obtaining a lower value of m = m2.The corresponding values of the gradients m,and m2 for the five clays (with sufficient data)are recorded in Table 1; the Boston blue claywas reconstituted, the other four undisturbed.

More attention deserves to be paid to various insitu geophysical tests. Pile load tests should alsobe considered to be a special form of in situtests; more information could be obtained aboutthe elastic properties of the ground by carefulobservations during the unloading of the testpile.

The compilers of the table have given theirviews of the applicability of each test for deduc-ing a variety of soil properties. The two instru-ments that have the best ratings by far are theself-boring pressuremeter and the piezocone(with friction sleeve). The interpretation of typi-cal results from each instrument is discussed inthe following sections.THE SELF-BORING PRESSUREMETER

The importance of the relationship in equa-tion (29) is that if the undrained strength ratio ofa clay can be measured or estimated for a nor-mally consolidated specimen then its value canbe predicted for other specimens of the clayprovided that the degree of overconsolidation isknown. This point is returned to later.RANGE OF IN SITU TESTS

The pressuremeter was first conceived byMCnard in 1954, as a tool which was insertedinto a preformed borehole and used for themeasurement of the strength and stiffness ofsoils and rocks. Since MCnards pioneering workthere have been many major developments inthe pressuremeter, especially in the last 15years, which can be conveniently divided intofour categories:

A good indication of the wide variety of in (a) techniques of measurementsitu tests that have been developed is provided (b) techniques of insertionby Table 2. This table is taken from a report by (c) new analyses of the expansion testCampanella & Robertson (1983) which forms (d ) new types of test.

459


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460 INTERPRETATION OF IN SITU SOIL TESTSTable 2. Perceived applicability of in situ test methods-update 1982 (ufter Campsnella & Robertson, 1983)*

Dynamic cone C A B C C - c - - - -Static coneMechanical B A B C B - C B c--Electrical friction B A B C B - C B c--Electrical piezo A A B B B A A B B A BElectrical piezo/friction A A A B B A A B B A B

Acoustic probe C B B C C - c c - - -Dilatometer B A B C B - B B c--Vane shear B C - - A - B - - - -Standard penetration test B B B C C - - _ c--Seismic cone penetration

test downhole C C C - - - _ A - - -K, blade - - - - - - B - - - -Resistivity probe B B A B C - c c c--Borehole permeability C - - - - A - - - B AHydraulic fracture - - - - - B B - - c cScrew plate C C B C B - B A B C CSeismic downhole C C C - - - - A - - -Impact cone C B C C C - c c c--Borehole shear C C - B B - c c - - -Mtnard pressuremeter B B C B B - C B B--Self-boring pressuremeter B B A A A A A A A A BSelf-boring devices

K,, meter - - - - A - - - -Lateral penetrometer C C B B B - B B c--Shear vane B C - - A - B - - - -

Seismic cross-hole C C B - - - _ A - - -Nuclear tests - - A B - - _ CPlate load tests C C B B C - B A B C C

- C- B- BB AB A- CC B- AB B- _C A- _- -B BB B- Cc -c cA A- _- _B B- CB B

* A, high applicability: B, moderate applicability; C, limi [ted applicability.

In the basic expansion test of the pressureme-ter, measurements are made of the pressure andof the increase in volume of the inflatablemembrane. In the early equipment used byMenard these observations were made at theground surface by monitoring the pressure andvolume of the fluid used for inflation. With thedevelopment of accurate and reliable electricaltransducers placed within the pressuremeter ithas been possible to improve the accuracy ofthese measurements by an order of magnitude

or more. For example the feeler gauges incorpo-rated in the Cambridge self-boring pressureme-ter are sensitive to movements of the order of0.005 mm, which is equivalent to less than0.02% volumetric strain for an instrument witha diameter of 83mm (as in the latest version).

The self-boring pressuremeter has been de-veloped in parallel in France and Britain, wheremost of the fundamental work has been done atthe University of Cambridge by a succession ofresearch students and fully reported in their


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WROTH 461PhD theses: Hughes (1973), Windle (1976),Clarke (1981) and Fahey (1980).

It is presumed that the purpose and details ofthe self-boring technique are well understood.To minimize the disturbance caused by inser-tion, various studies have been carried out of thedegree of disturbance created by differentmodes of operation and design of equipment.The geometrical arrangement of the cutting shoeand cutting tool is crucial. Recent work byClarke (1981) suggests why the push-in pres-suremeter may prove to be a successful com-promise for offshore use between the complexself-boring pressuremeter and the cruderMCnard instrument.Initially, the results of pressuremeter testswere interpreted by means of empirical expres-sions to give parameters for design such as al-lowable bearing capacity factors and moduli forallowable settlement. The first fundamental in-terpretation of an expansion test was suppliedby Gibson & Anderson (1961) in which thepressuremeter was considered to be infinitelylong so that the deformation of the surroundingsoil was assumed to be in conditions of axialsymmetry and plane strain. One analysis wasapplicable to undrained expansion tests in clayin which the soil is assumed to undergo zerovolume change and to behave as an elastic-perfectly plastic material characterized by ashear modulus G and an undrained shearstrength s,. A separate analysis was achieved forexpansion tests in cohesionless soils for whichthe soil is assumed to behave elastically untilfailure occurs at a constant effective stress ratio(governed by the Mohr-Coulomb criterion interms of a friction angle 4) and with no volumechange.Both of these analyses have been improved.The undrained expansion test can now be in-terpreted with no prerequisite assumption ofstress-strain properties of the soil being tested,as a result of work independently done byBaguelin, JCzCquel, Le MCe & Le Mehaute(1972), Ladanyi (1972) and Palmer (1972). Un-fortunately the interpretation is very sensitive tothe disturbance of the soil caused by insertionand to the datum selected for the strain.

The second case, the drained expansion test ina cohesionless material, has been modified byHughes, Wroth & Windle (1977) to take ac-count of the volume change that occurs afterfailure has been initiated. The relevance of thisnew analysis has been confirmed by Fahey(1980) by a special series of laboratory testsunder carefully controlled conditions.

To date most pressuremeter tests have con-sisted of a rapid expansion test either conducted

with stress increments applied at regular inter-vals or under constant rates of strain. With theadvent of microprocessors and computer-controlled experiments, as well as additionalobservations, e.g. pore pressure measurements,it is now possible to carry out new types of test.One such possibility is a holding test formeasuring in situ the consolidation characteris-tics of the soil being tested. Clarke, Carter &Wroth (1979) report the results of such holdingtests in which a quick undrained expansion testin a clay to, say, 10% strain is followed by aphase when the membrane is kept automaticallyat its (enlarged) radius by adjustment of theapplied pressure. During this phase the clayundergoes consolidation with a reduction in thetotal radial stress and excess pore pressure, bothof which are continuously recorded. The coeffi-cient of consolidation can be deduced from theresults.Good quality pressuremeter test in c lay

In this and the next section examples aregiven from one site of two good quality pres-suremeter tests, one in clay and one in sand. Thesite is part of the outer harbour at Zeebruggewhere it was planned to construct large storagetanks for liquefied natural gas. As part of theoverall site investigation, PM Insitu TechniquesLtd were commissioned by Distrigas NV tocarry out a profile of self-boring pressuremetertests; the work was supervised by Tractebel Zacting as engineers.

The soil profile at this site consists ofhydraulic-pumped fill to a depth of 15 m,quaternary formations consisting mainly ofsands from 15 m to 33m, Bartoon clay from33 m to 46 m and below this Pamiselian de-posits, predominantly sandy soils.

The self-boring pressuremeter cannot be dril-led through soil containing hard inclusions, suchas sandstone layers and gravel which occur atthis site. It cannot be drilled continuouslythrough sand deposits as it leaves an unsup-ported borehole which may collapse and makeextraction of the instrument difficult if not im-possible.

On this site it was therefore necessary to use adrilling rig to provide a borehole which wouldnot collapse and to cope with hard inclusions.The borehole was drilled and cased using a shelland auger rig to a depth of 1 m above the levelof a proposed test position. The self-boringpressuremeter was lowered down the boreholeand drilled beyond the bottom of the casing toat least 1 m into material presumed to be undis-turbed. A pressuremeter test was carried out. Incertain instances it was possible to advance the


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462 INTERPRE TATION OF IN SITU SOIL TESTSself-boring pressuremeter 1 m further and tocarry out a second test. The self-boring pres-suremeter was withdrawn from the borehole sothat the hole could be drilled further by the shelland auger rig. The one borehole was terminatedat a depth of 88 m and included 24 pressureme-ter tests.

The results of a pressuremeter test in clay at adepth of 43.4 m are reproduced in Fig. 10. Thereadings are automatically recorded on site bothon magnetic disc and as a hard copy on paper,and subsequently processed and plotted by com-puter. The pressure 9 applied to the inside ofthe membrane is plotted against the cavity straine which is the observed expansion of the mem-brane divided by its initial radius a,. The pres-sure 4 has been corrected for the (small)strength of the membrane itself, represented bythe difference between the ordinates of points 0and I.

At the start of the expansion test the mem-brane fits tightly over the body of the instrumentand has the same diameter as the cutting shoe.In theory no expansion of the membrane shouldbe detected until the applied pressure $ is equalto the in situ total lateral stress in the ground incontact with the pressuremeter. In reality therewill be some small compliance of the instrumentitself, until at point A (the lift-off pressure), onthe expansion curve, the soil starts to deformunder increasing lateral stress.It is usually assumed (both for convenience

zooor F

I-2 0 2 4 6 a 10 12D %

Fig. 10. Resufts of a pressuremeter teat in Bartoonday, Zeebrugge

and for lack of better information) that thelateral stress in the ground is the same in allhorizontal directions; this is unlikely to be thecase except in very homogeneous deposits sub-ject solely to one-dimensional consolidationduring their geological history.The Cambridge self-boring pressuremeter hasthree separate feeler arms for measurement ofthe movement of the membrane, which are sym-metrically disposed at 120 around the section ofthe instrument. By careful scrutiny at suitablylarge scale of separate plots of $ against each ofthe values of F, three separate estimates of thein situ total lateral stress can be obtained (on thepresumption that there has been negligible dis-turbance of the ground during insertion of thepressuremeter). The results in Fig. 10 are forone of the three feeler arms and indicate a valueof oh0 of 646 kN/m. The measurement of in situlateral stress by the self-boring pressuremeter isdiscussed in detail by Ghionna, Jamiolkowski &Lancellotta (1982) and by Lacasse & Lunne(1982a).

In the original analysis of Gibson & Anderson(1961) for the interpretation of an undrainedpressuremeter test in clay, it was assumed thatthe clay behaves as a perfectly elastic-perfectlyplastic material characterized by a shear mod-ulus G, Poissons ratio v = 0.5 and an un-drained shear strength s,. The analysis leads tothe result that, after failure has been initiated inthe clay, the applied pressure 4 is linearly re-lated to the logarithm of the c ur rent volumetricstrain AVIV. Moreover, if natural logarithms areused the gradient of the line will be equal to theundrained shear strength, denoted here by sUD,.

The field results of the pressuremeter test ofFig. 10 have been replotted in Fig. 11 in the

/2000r

///

I 7 8 910 12 14 16 18 20ZVIV o/o

Fig. 11. Processed results of a pressuremeter test inBartooo day


15/41

WROTH 463above manner as J, against log(AV/V). Notethat the volumetric strain is directly related tothe observed strain by the relationship

AV/V=l-(l+~))~ (31)The points D, E and F correspond to those

selected in Fig. 10. The results lie remarkablyclose to a straight line, and that selected has agradient s,,,= 386 kN/m. The newer andsuperior analysis of Baguelin et al. (1972),Ladanyi (1972) and Palmer (1972) does notrequire an assumption to be made about thestress-strain behaviour of the soil. It leads to theresult that the local gradient of the curve drawnthrough the points in Fig. 11 is the shear stresscorresponding to the shear strain experienced atthat stage of the test by the clay in contact withthe pressuremeter. The fact that the curve is soclose to a straight line vindicates the assumptionimplicit in the Gibson-Anderson analysis thatthe clay behaves as a perfectly elastic-perfectlyplastic material.

From experience gained at a large number ofsites it is believed that the Gibson-Andersonanalysis is satisfactory for design in geotechnicalengineering and that the extra sophistication ofthe 1972 analysis is not warranted, particularlybecause it is so sensitive to the choice of datum.General experience from a large number of sitesindicates that the in situ undrained shearstrength obtained in pressuremeter tests is con-sistently greater than corresponding values fromthe vane shear test, cone penetrometer test andplate bearing test. This finding is discussed byMair & Wood (1984) who cite examples quotedby Hughes, Wroth & Pender (1975), Windle &Wroth (1977) and Ghionna, Jamiolkowski,Lacasse, Lancellotta & Lunne (1983).

The shear modulus of the clay can be meas-ured by arranging for an unloading-reloadingcycle such as BCB to be included in the expan-sion test. If the clay behaves as a perfectlyelastic material in unloading then BC will be astraight line of gradient 2G (using small straintheory). It should be noted that the cycle BCB ismarkedly linear with very small hysteresis, andgives a value of the shear modulus G =47 MN/m.In carrying out such an exercise care must betaken not to exceed the elastic limit of the clayduring the unloading phase; this restricts theamplitude of the stress cycle that can be applied.For the ideal elastic-plastic material the allowa-ble amplitude is twice the undrained shearstrength, as shown by Wroth (1982).

It is therefore possible to construct a curvesuch as JKG in Fig. 10, which serves as a

700i

600.

I- 2 0 2 4 6 8 10 12F Q/ o

Fig. 12. Results of a pressuremeter test in sand at

theoretical boundary to the elastic behaviour inunloading. It should be noted that this boundaryfits well with the extent of the linear portion ofthe final unloading part of the test FGHI.Good quality pressuremeter t est n sand

The results of a pressuremeter test in sand (inthe same borehole as that for the test in clay) ata depth of 10.6 m are reproduced in Fig. 12.The shape of the curve is different from that ofthe test in clay of Fig. 10, both in the loadingportion MPQRT and in the final unloading por-tion TUVW. This difference is always observedin good quality tests and can be used as anindicator of the material tested. It can be rep-resented numerically by the factor 0 defined as(Baguelin, 1982)

B=p20-p5Pm- PO (32)where po, pS and pzo are respectively the appliedpressure at O%, 5% and 20% volumetric strain.Noting that the volumetric strain is approxi-mately twice the plotted strain E, the value of /3for the clay in Fig. 10 is 0.45, whereas that forthe sand in Fig. 12 is 0.51.

The same method as that outlined previouslycan be used for an assessment of the in situlateral total stress. When the results of Fig. 12are plotted to a larger scale, the lift-off pressureis given by the point L, i.e. a stress of107 kN/m*.


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464 INTERPR ETATION OF IN SITU SOIL TESTS

Fig. 13. Processed results of a pressuremetersand at Zeebrugge

test in

In the test on a cohesionless material which issufficiently permeable, the final part of the un-loading curve VW provides valuable informationabout ground conditions. As the membrane isdeflated, ultra-loose sand collapses into the en-larged borehole under very small effective stres-ses. In the extreme condition at W, the mem-brane is being pushed back to its initial size bythe action of the groundwater, and the pressurerecorded inside the membrane is a measure ofthe ambient pore pressure u0 given by the hori-zontal tangent WVZ. (Note that point W is thelast one automatically recorded during the test,and the computer when producing the plot ofFig. 12 has inappropriately drawn the chord WOinstead of continuing the line VW back to zerostrain.)The value of u. is 62 kN/m* corresponding toa water-table at a depth of 4.3 m. The water-table in the hydraulic till was observed tofluctuate between wide limits as the tide levelvaried. With the estimate that the total verticalstress, at the test depth of 10.6 m, is 160 kN/m,this suggests a value of the coefficient of earthpressure at rest, Ko, to be (107 - 62)/(160 -62) = 0.46.In the analysis of Hughes, Wroth & Windle(1977) for the interpretation of a drained pres-suremeter test in sand, it is assumed that thesand behaves elastically before failure and failsat a constant ratio of effective stresses and at aconstant rate of dilation. The analysis leads tothe result that, after failure has been initiated in

the sand, the logarithm of the effective radialstress 4 - u0 is linearly related to the logarithmof the strain E; the gradient of the line s is givenby the expression

sin $(I +sin V).S= 1 +sin 6 (33)

where 4 is the angle of internal friction and v isthe angle of dilation.

The field results of the pressuremeter test ofFig. 12 for the range PQR have been plotted inthis manner in Fig. 13. They approximate to alinear relationship whose gradient s = 0.425.The adoption of a value of 35 for the criticalstate angle of friction at constant volume gives avalue of 4 = 39 and v = 9.5 (see Hughes et al.,1977).The shear modulus of the sand can be meas-ured in the same way as that for the clay, bycarrying out small unloading-reloading cyclessuch as MN and RS in Fig. 12. The correspond-ing value of the shear modulus for cycle MN is3 1 MN/m.As for the clay there is a limit to the elasticrange of behaviour during unloading. For theidealized sand behaviour used in the analysis,the allowable amplitude of the stress cycle QY is

as illustrated in Fig. 12 (Wroth, 1982).It is therefore possible to construct a curve,such as XYU in Fig. 12, which serves as atheoretical boundary to the elastic behaviour inunloading. This has been based on the observedvalue of 4= 39 to give ratios of 0.772 and0.228 used in the plot. It should be noted that(T, varies with the stage of the test so that thedistance QY would not be the same if Q were tobe chosen elsewhere on the virgin loading curve;care must be taken to subtract the ambient porepressure L+, rom the applied pressure J, to givethe relevant value of (T, acting on the sandwhich is in contact with the membrane. Thetheoretical boundary fits well with the extent ofthe linear portion of the final unloading part ofthe test TUVW.Correlation of undrained shear strength

The interpretation of undrained shearstrength from pressuremeter tests in terms ofeffective stresses is uncertain with the presentstate of knowledge. This is because there havebeen very few attempts to test soil under therelevant combination of consolidation and sub-sequent shearing.


17/41

WROTH 465To study the pressuremeter expansion test in

the laboratory by testing single elements of soil,it is necessary to be able to apply to the elementthe strain path to which the soil elements in thefield are subjected. This strain path involvesone-dimensional consolidation followed byshearing under conditions of plane strain, atconstant volume, in the plane perpendicular tothe direction of consolidation. True triaxial de-vices are the only types of apparatus that canapply this complete strain path in one continu-ous operation without the need for unloading,trimming and reorientating the sample.

A limited number of such tests has been re-ported by Wood & Wroth (1977), Wood (1981)and Eden & Law (1980); the results have beencompared with those computed from variousgeneralized stress-strain models of soil be-haviour, none of which proved to be satisfac-tory.

For example, one approach is to use the mod-ified Cam clay model, adapted for plane strainconditions, and to derive for the undrainedstrength ratio for anisotropically normally con-solidated specimens (see Appendix 2)

SUP, sin I$,, c2+ 1 A-=- -flO ( 1c 2 (34)

where c = l/(2-sin &,.). This expression fitsdata of conventional plane strain active testssatisfactorily, e.g. for Boston blue clay, takingI#+,= 30 gives &= 33.75, c = 0.6923 ands,psloVo)= 0.315 which compares with an aver-age value of O-338 from the results reported byLadd & Edgers (1972). However, if this expres-sion is applied to the simulated pressuremetertests conducted in the true triaxial apparatus onkaolin by Wood then the predictions are notsatisfactory. Wood & Wroth (1977) quote avalue of &,, = 26 which corresponds to c =0.640 and s,JuVO = 0.259, whereas the ob-served value of sUpS/uV,, as as low as 0.179.Clearly the use of the plane strain version of themodified Cam clay model is inappropriate formodelling the data from pressuremeter tests.

Apart from a lack of success in the mathemati-cal modelling of those laboratory tests whichsimulate pressuremeter tests, there are impor-tant differences between the single-element testsconducted under ideal conditions in the laborat-ory and pressuremeter tests carried out in thefield.

The most important difference is that in theground around the pressuremeter the fields ofstress and strain do not remain homogeneousduring an expansion test, with the consequencethat there are high gradients of excess pore

pressure in the radial direction, so that somepartial consolidation will occur in a supposedlyundrained test. To minimize this effect, mostpressuremeter tests are conducted quickly and atstrain rates much faster than those normallyused in conventional laboratory tests. This, inturn, introduces another undesirable effect inthe influence of strain rate on undrained shearstrength due solely to increased viscosity whichis entirely separate from consolidation effects.Because of the inhomogeneity of stress andstrain, the various annuli of soil arounu a pres-suremeter experience different rates of strain, sothat it is not possible to make a direct allowancefor the higher strain rates used in the field.

The arguments about partial consolidation ina pressuremeter test can be illustrated qualita-tively in Fig. 14 by considering what wouldhappen in a test on an idealized soil which iselastic-plastic but obeys the Mohr-Coulombfailure criterion. Fig. 14(a) represents a sec-tion of the ground perpendicular to the axis ofthe pressuremeter showing the radial distribu-tion of excess pore pressure given by the curveFH assuming that no consolidation occurs. Figs14(b) and 14(c) show the effective and totalstress paths, and the stress-strain response ofelements of soil D, E, F and G at different radialdistances from the pressuremeter. The pointswith s&ix 1 relate to theoretical stress stateswith no consolidation, whereas those with suflix2 relate to the more realistic situation in whichsome consolidation has taken place. The impor-tant point is that the partial consolidation meansthat the effective stress states Dz and EZ moveup the effective stress failure envelope, and thecorresponding soil elements are failing undercontinually increasing shear stress, i.e. the in-stantaneous value of the so-called undrainedshear strength is increasing monotonicallythroughout the pressuremeter test.

Although the precise details of Fig. 14 are notcorrect, the principles it contains apply to a realtest on a real soil. In the early days of thedevelopment of the self-boring pressuremeter ananalysis was attempted by Wroth & Hughes(1972) to interpret a slow expansion test in clayin which full consolidation would occur.There is no doubt that current practice in theinterpretation of rapid pressuremeter tests inclay, which does not allow for the effects eitherof partial consolidation or of strain rate, leads tooverestimates of the undrained shear strength ofthe undisturbed clay. At present it is not possibleto know whether the overestimates are by a mar-gin of lo%, 20% or even 50%, and the marginmust be a function of the coefficient of consoli-dation of the clay in question. There is an urgent


18/41


(b) (C lFig. 14. Changes in the total and effective stress states around a pressuremeter due to partialconsolidation during an m&a&d test

SIPS near Grangemouth Test LB1 Depth 1.90 m

xgh= 41 kN/m2S = 60 kN/m*G = 2900 kN/m*

(b)I

q = 14 m2/year

0 20 40 60 80CC ) Tfme mr

Fig. 15. Typical holding test results (after Clarke etal., 1979)

need for research to be carried out to resolvethese uncertainties.Holding test

A typical set of results from a holding testcarried out in soft clay at a site nearGrangemouth is shown in Fig. 15 in which theobserved quantities E, Q and Au are plottedagainst time, where Au is the excess pore pres-sure measured at the membrane-soil interface.

The initial part of the holding test is carriedout in the same manner as an undrained pres-suremeter test with a rate of cavity strain ofabout 1% per minute. By using the method ofinterpretation already outlined the values of who,s, and G can be determined. As well as beingimportant in its own right for design, the valueof s, is necessary for the interpretation of thesubsequent consolidation data and the rigidityindex G/s, allows a check to be made on theexcess pore pressure generated during the ex-pansion phase of the test.

At about 9.5% strain the rate of straining isgradually reduced to zero over about 30 s. Themembrane is then held fixed at this inflatedradius until completion of the test by automaticreduction of the internal pressure 9 to matchthe reduction in external total radial stress asconsolidation occurs in the surrounding soil.Careful monitoring of the decay of pore pres-


19/41

WROTH 467sure allows an estimate to be made of the con-solidation characteristics of the soil.To interpret the data from a holding test amathematical solution is required for this par-ticular boundary value problem. In addition tothe assumptions for an undrained expansion testin clay, it is assumed that during consolidationmovements of the soil skeleton and flow ofporewater will be entirely radial.

It has been shown (Gibson & Anderson,1961) that the maximum excess pore pressureafter undrained cavity expansion occurs at themembrane-soil interface and has a value

Au,, = (35)

Further, it has been shown that the distributionof excess pore pressure, immediately after ex-pansion, is logarithmic with radius within thezone of yielded soil and is zero in the outerelastic region.Numerical solutions for the consolidation ofan elastic, perfectly plastic soil around an ex-panded cylindrical cavity have been obtained(Carter, Randolph & Wroth, 1979). These indi-cate that the total membrane pressure will de-crease gradually once membrane expansionstops and as the soil consolidates. This feature isalso observed in the field, e.g. see Fig. 15(b).In principle, it would be possible to use therate of decay of this total pressure to provide ameasure of the horizontal coefficient of consoli-dation of the soil, c,,. However, the magnitudeof this pressure drop is not usually large and amore accurate measure of c,, can be obtained bymonitoring the dissipation of the excess porepressure at the membrane-soil interface.

A closed form solution for the time depen-dence of the excess pore pressures around dri-ven piles, but also relevant to the pressuremeterproblem, has been found by Randolph & Wroth(1979). They assumed that the soil behaves en-tirely elastically during consolidation. Subse-quently, it has been shown that the solution foran elastic soil is sufficiently accurate for anelasto-plastic soil (Carter, Randolph & Wroth,1979).In Fig. 16 solutions for the time for 50%consolidation have been plotted against themagnitude of the maximum excess pore pressureAu,,, (normalized by s,). The non-dimensionaltime factor T,, has been used, where

Tso = chUr,z (36)and fXO s the real-time interval required for thepore pressure at the membrane to reduce to halfits maximum value, r,,, is the radius of the

:: /Oh -4 -3 -2 -1 0 1 2

T5oFig. 16. Tie for 50% pore pressure decay at themembranc+soil interface (after Randolph & Wrath,1979)

membrane after expansion and c,, is the horizon-tal coefficient of consolidation.The coefficient c,, is related to other soil prop-

erties by the equationk, 1-vc,=-2G--Y.JJ l-2v (37)

where k, is the horizontal permeability of thesoil, yw is the unit weight of porewater and v isthe drained value of Poissons ratio.

It is possible to fit the theoretical solution forconsolidation to the experimental curve of ex-cess pore pressure versus time at the 50% level.This is directly analogous to the standardlaboratory procedure of fitting Terzaghis solu-tion for one-dimensional consolidation tooedometer results. For example, in Fig. 15(c)the maximum excess pore pressure Au,,, is ap-proximately 120 kN/m* and occurs 10 min afterthe start of testing. This excess pore pressure hasdropped to half its maximum value 25 min afterthe start. Hence the time tsO is given by f5,,=25 - 10 = 15 min. From the expansion portion ofthis test a value of the undrained shear strengths, = 60 kN/m is deduced in the usual manner(Gibson & Anderson, 1961). Hence Au,,,/s, = 2and a corresponding value of T,,,=O.33 can beread from Fig. 16. The final membrane radius inthis test was r, = 0.0352 m. Thus

0.33 x 0.0352 x 0.035215

= 2.7 x lo- m*/min= 14 m/year

Clarke et al. (1979) quote results of holdingtests conducted at Canvey Island and make acomparison between values of the coefficient of


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468 INTERPRETA TION OF IN SITU SOIL TESTS

Water pressure: Friction:MN/m MN/m21.0 0.5 0 0.2 0.1r I

-3oL

FrictionCone resistance: ratio SOllMN/m2 f& x 100 profile4 8 12 16 20I

2 Predril ledb

Fig. 17. Results of an onshore piezoco ne test (after Zuidberg et al., 1982)consolidation c,, deduced from holding tests withvalues of c, obtained from conventional oedo-meter tests; the former values range from afactor of 30 to 300 greater than the latter val-ues. The field values are much more in line withvalues back calculated from settlement recordsof engineering structures.

The coefficient of consolidation, or permeabil-ity, is an instance of an important soil propertywhere values measured in the laboratory may beseriously misleading and give rise to gross over-estimates for times of settlement. It is suggestedthat the holding test is an important type ofpressuremeter test that should be developed.Apart from allowing realistic estimates of con-solidation characteristics to be made, it mayprove crucial in the interpretation of undrainedshear strength to allow for effects of partialconsolidation, as discussed in the previous sec-tion.

During the past 15 years the self-boring pres-suremeter has been developed to the stagewhere it has become established as a major toolfor in situ measurements of the following soilproperties: in situ lateral total stress, shearstrength, shear modulus and consolidationcharacteristics. The range of soils and soft rocksin which it can be used has been extended toinclude boulder clay, loose, dense and weaklycemented sands, and soft rocks such as mud-stone, siltstone and chalk.

THE PIEZOCONE TEST

The basic cone penetrometer test has beenused for many years as a standard tool in siteinvestigation especially for establishing quicklyand cheaply the soil profile at a particular site. Ithas grown in importance and acceptance withthe recent special needs of offshore site investig-ation associated with oil exploration.

The first attempts to record porewater pres-sures during penetration of a probe into theground were reported separately by Torstensson(1975) and Wissa, Martin & Garlanger (1975).This pioneering work showed the clear promiseof incorporating a pore pressure transducer in acone penetrometer so that continuous profiles ofpore pressure could be obtained at the sametime as point resistance of the cone and meas-urement of sleeve friction. The resulting instru-ment has become known as the piezocone.

The development since then has been rapid,with a variety of instruments on the market withmost applications being offshore. Nearly all in-struments have been built to the standard di-mensions of the Dutch cone (60 apex angle andbase area 10 cm) but unfortunately the positionof the pore pressure transducer has not beenstandardized. The position has a marked in-fluence on the magnitude of the observations, asdiscussed later in this section.

Figure 17 shows a typical profile of piezocone


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WROTH 469

compression tests

results at an onshore site in the Netherlandsreported by Zuidberg, Schaap & Beringen(1982). The soil profile was obtained from aninterpretation of the cone resistance, sleeve fric-tion and friction ratio. In the standard Fugrocone the pore pressure transducer is locatedhalf-way up the conical tip. The profile of thepore pressure provides dramatic confirmation ofthe soil profile. The hydrostatic conditions areshown by the chain-dotted line, so that the offsetbetween that line and the continuous recordedprofile represents the excess pore pressure. Dur-ing penetration of the relatively permeable sandthe excess pore pressures are small but justpositive, whereas the excess pore pressures re-corded in the lower stratum of clay are large andpositive. The magnitude of the excess pore pres-sure is an indication of the type of soil and forclays can be correlated in dimensionless formwith the overconsolidation ratio. The sharp fluc-tuations in the pore pressure act as a sensitiveindicator of local variations in the soil such aslaminations or fine partings of silt. In specialcircumstances of a dense cohesionless soil thatexhibits dilatancy when sheared the excess porepressures may be negative; an example of aweakly cemented sand is given by Ventura(1983) and of a silt (after dynamic compaction)by Campanella et al. (1982).

Confirmation of the type of soil can be ob-tained by stopping penetration of the cone andobserving the dissipation of the pore pressure asconsolidation occurs, in exactly the same way asa holding test with a pressuremeter.It is evident that the piezocone provides im-portant data about soil conditions, but at presentthis is of a qualitative nature. To advance itspotential it is necessary to be able to interpretthe data in a reliable and consistent quantitativemanner. This requires both a standardization ofequipment, test procedures and interpretation aswell as a clear understanding of soil behaviourand the excess pore pressures generated duringshear. To illustrate some essential points refer-ence is made in the next section to basic soil

behaviour as observed in triaxial tests. It issuggested that this pattern of behaviour can beused as a framework for the interpretation ofdata from the piezocone.Interpretation of pore pressures in the triaxial test

Consider a saturated specimen of clay that hasbeen isotropically normally consolidated to thestress state represented by point A in Fig. 18. Ifa conventional undrained triaxial compressiontest is carried out the effective stress path will bea curve such as AB, with the specimen reachingthe critical state at B (see for example the dataof kaolin in Fig. 6). The total stress path will bethe straight line AD which is necessarily ofgradient 3 in the chosen stress space. At failurethe excess pore pressure generated within thespecimen is represented by the difference be-tween the mean total and effective pressures, i.e.the distance BD. This observed magnitude ofexcess pore pressure is made up of two compo-nents: the component BC due to the response ofthe clay to the shearing process represented byAq and the component CD due to the change inmean total stress Ap applied to the specimen.

If instead an unconventional undrained com-pression test had been conducted on the speci-men such as one with the total stress path AE(achieved by increasing the cell pressure approp-riately as the axial stress is increased) the effec-tive stress path would have remained unchangedas AB. However, the observed magnitude of theexcess pore pressure BE would be greater thanthe value BD measured in the conventional test.This difference cannot reflect any difference insoil behaviour, but merely provides evidence ofthe different changes in mean total stress Ap inthe two tests selected by the operator.

Hence the magnitude of the excess pore pres-sure measured in a shear test is not a uniqueproperty of the soil behaviour but it dependsalso on the changes in total stress applied exter-nally to the specimen. The first component BC isa unique property of the soil (when tested intriaxial compression) and can be correlated with


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470 INT E RPRE T AT ION OF IN S IT U SOIL T E ST S

10 608-

06- l

_/I

Weald clay

04 - l

d 0.2.

O-

l .

-0-61 1.5 2 3 456,a 810 1520I 3040 1Overconsol~datmn ram

Fig. 19. Variation in the pore pressure parameter A atfaffnre with the overconsolidation atio for W eald clay(after Bishop & Henkel, 1957)other properties of the soil specimen such as itsoverconsolidation ratio. The second componentCD or CE gives no information whatsoeverabout the soil being tested but is just the valueof Ap experienced by the specimen.

Consequently it is vital in interpreting obser-vations of pore pressure changes to make dueallowance for the change in Ap which accom-panies the application of shear stress. This dis-tinction is normally made in the interpretationof pore pressures observed in triaxial tests.For a saturated clay the pore pressureparameters originally introduced by Skempton(1954) are such that for triaxial compressiontests

A= Au - Aa3Aa1 -bug (38)and B is taken as unity. Essentially the parameterA is a ratio of the pore pressure change to thechange in deviator stress, in which allowance hasbeen made in the numerator for any change incell pressure Au,, but not unfortunately for anychange in mean total stress hp. This means thatthe value of A depends on the test conditions aswell as soil properties such as the overconsolida-tion ratio.

To meet this objection Henkel (1960)suggested a revised set of parameters defined sothat for a saturated specimen

Au - Au,,,a= Ar0ct (39)

where Au,,, =Ap is the change in octahedralnormal stress (or mean total stress) and AT,* is

the change in octahedral shear stress defined as7,t = $[(a2 - a$+ (u3 - a,) + (a, - u2)2112

(40)The concepts of CSSM allow a relationshipbetween A and OCR to be derived as follows.

In Fig. 5 the isotropically overconsolidatedspecimen initially at state R reaches the criticalstate at S. For a conventional triaxial compres-sion test in which the cell pressure is kept con-stant, Au, -0 so that Ap = +Aq and the totalstress path is the straight line RT of gradient 3.The excess pore pressure is given by

Au=p,-p,= pr+Ap-pp,= Pr+fq.-P.

Hence the pore pressure parameter A is expres-sed as

A= Au-Au3Au, - Aa3AU=-4.

= P;+fq,-P,MP,

but using equation (20) this can be written asA=;[(;)-*+;-I] (41)

which is an expression relating A with the over-consolidation ratio R of the specimen and withthe soil properties A, M and r.

The variation in values of A at failure withOCR for remoulded Weald clay reported byBishop & Henkel (1957) is reproduced in Fig.19. The curve drawn by those authors throughthe experimental points is very closely matchedby the prediction of equation (41) with thefollowing values of the soil constants for Wealdclay: A =0.093 and K = 0.035 to give A=0.6237, M =0.95 and r =2, taken from Table6.1 of Schofield & Wroth (1968). The quality offit should be no surprise, as it is another way ofconfirming how well a remoulded low plasticityclay, such as Weald clay, tested in triaxial com-pression after isotropic consolidation satisfiesthe critical state concept.

The foregoing detailed analysis underlines theimportant relationship between the two dimen-sionless quantities, the pore pressure parameterA and the overconsolidation ratio. It suggests


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d

471

0 n5 10 1.5 2.0 2.5 3.0Normalrc?d excess pore pressure Au,/(Au)~,, Normaltzed exces s pore pressure AU,I(AU)~,~Fig. 20. Nonualized excess pore pressures along the face and shaft of 1s and 60 cones duringsteady penetration in Boston blue clay (after Baligh & Levadoux, 1980)

that A is linearly related to OCR raised to thepower of -A, and further that this can be usedas a basis for the interpretation of pore pressuredata from tests other than triaxial. This is at-tempted in the next section.Interpretation of pore pressures in the piezoconetest

The quantitative and detailed interpretationof the results of cone penetrometer tests has notyet been achieved because of the complex na-ture of the strain and stress changes induced inthe soil around a penetrating cone. The problemhas been studied by a number of workers, nota-bly Baligh & Levadoux (1980), and Fig. 20 istaken from their major report.

In this diagram experimental data are com-pared with computed results of the distributionof excess pore pressure around a piezocone. Theresults refer to two different geometries of coneshown in section in the centre of the diagram.The left-hand figure refers to a piezocone with acone of 18 apex angle and five separate trans-ducers at the positions indicated. The right-handfigure refers to a piezocone with a standard 60cone and two transducers.

The excess pore pressures Au occurring onthe boundary of the instrument have been nor-malized by dividing by the steady state valueAu,, that exists a distance up the shaft of theinstrument away from the tip; they are plotted

against the axial distance from the tip divided bythe shaft diameter.

The value of the excess pore pressure variesmarkedly with the position of the sensor inrelation to the cone. It is important that theposition of the sensor should be standardized forall piezocones to optimize the interpretation ofdata, particularly in view of the uncertaintiesinvolved and the unavoidable reliance on em-piricism to some extent.

It is tempting to suggest that the sensor shouldbe sited at the tip of the cone where the max-imum values of the excess pore pressure occur,but this is not the best position for a number ofreasons. Campanella et al. (1982) list severalsound practical reasons why the best location ison the shaft (at position such as No. 2 for the60 cone on the right-hand side of Fig. 20)rather than on the cone itself. There are threeadditional and major advantages from a theoreti-cal standpoint in favour of the location on theshaft.(a)

(b)

A greater proportion of the excess pore pres-sure is due to the component induced byshear compared with the component due tothe change in the mean total stress.The gradients of excess pore pressure withaxial distance are smaller, so there is greateraccuracy.(cl The values are less affected by any change in


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472 INTERPRETATION OF IN SlTU SOIL TESTS1.0

0.8L.63 .mu 0.4 :\z00.2

OC RFig. 21. Variation in the piezocone pore pressare ratiowith the overconsolidation ratio at Ons0y

the axial load experienced by the piezoconewhen the penetration is stopped either forthe next drill rod to be added or for adissipation test to be carried out.

On the basis of the understanding of soilbehaviour provided by the long history of ex-perience obtained from high quality triaxialtests, the ideal interpretation of piezocone datawould be to derive a pore pressure parameterexactly analogous to Henkels parameter and toexpect this to correlate closely with the overcon-solidation ratio, i.e.

Au - Au,,,a= Ar=oct = f( OCR) (42)Unfortunately there is no means of knowing thechanges in octahedral normal and shear stresses;indeed, the changes vary spatially around thepiezocone and depend on the unknown proper-ties of the soil which is being penetrated.

In place of a, a dimensionless parameter isrequired which must be a ratio of excess porepressure to some measure of shear stress.Robertson & Campanella (1983) discuss theratios that have been used including(a) u/q, (Baligh, Azzouz, Wissa, Martin & Mor-

rison, 1981: Tumay, Bogges & Acar, 1981)(b) Au/q, (Campanella & Robertson, 1981)(c) Au/(q,-u,) (Smits, 1982)(d) Au/(q,-uVO) (Senesset, Janbu & Svana.1982; Jones & Rust, 1982; Jefferies &

Funegard, 1983)where qc is the measured cone end bearingpressure and qt is the corrected end bearingallowing for the effects of pore pressure actingon the back of the cone. Robertson & Cam-

panella (1983) rightly point out the importanceof making this correction, especially as it willvary markedly with the precise mechanical de-sign of the piezocone, and with depth below thewater-table.

The first three ratios are all unacceptable, thefirst because it has the total not the excess porepressure in the numerator, and all three becausethe denominators cannot be a proper measure ofthe shear stress. In an undrained situation themaximum shear stress can only be expressed interms of principal stresses as a difference of twototal stresses or a difference of two effectivestresses. It is for this reason that the variableqc - uVOand not qc is used for deriving values ofundrained shear strength from cone penetrationtests.

The last ratio, Au/(q,-a,,,), denoted B, bySenneset et al. (1982), is the one recommended,therefore, for adoption as the standard parame-ter for interpretation of piezocone data. It isassumed that proper calibration and correctionshave been made so that q= here replaces q,.

There are few reported instances of siteswhere piezocone data can be correlated withreliable data of OCR. One of the best re-searched sites is that at On&y used by theNorwegian Geotechnical Institute. From the in-formation reported by Lacasse & Lunne (1982b)values of B, have been calculated from theplotted profiles of Au, q= and mVo or depths atintervals of 2 m up to 20 m. These have beenplotted in Fig. 21 against the relevant value ofOCR taken from the profiles resulting fromconventional oedometer tests. The data aregiven in Table 3.

There is a striking similarity between the re-sults of the piezocone data presented in Fig. 21and the results of triaxial data in Fig. 19. In-deed, the theoretical background to the triaxialtest suggests that B, might vary linearly withOCR raised to the power -A, in a mannerTable 3. Processed data from piezocone tests fromOns0y

4,: Au:kN/m kN/m2200 56240 84


25/41

WROTH 473analogous to equation (41). If a value for A wereknown, then such a plot could be made with anexpected linear relationship. In practice, fielddata will not warrant such involved processing,because the variation in OCR in any onestratum is not likely to be great and because ofthe likely experimental scatter in the data. Aplot similar to Fig. 21 should be adequate forcorrelation.

It has been argued in this section that toderive the maximum benefit from the develop-ment of the piezocone(a) the position of the pore pressure transducer

should be standardized and should be on theshaft of the instrument and not within theconical tip

(b) the interpretation of the data should bestandardized by use of the pore pressureparameter B, = Au/(q,- aVo) in which qc isthe correct total end bearing pressure of thecone.

No attempt is made to consider the derivationof values of undrained shear strength from endbearing data of cone penetrometer tests. Atpresent the interpretation is usually empirical,being based on comparisons with other means ofmeasurement of s,. Much current research isaimed at a properly formulated method of in-terpretation, such as that of Senesset et al .(1982).DIRECT SIMPLE SHEAR TESTS

It was pointed out earlier that there will bemajor differences between tests in which theprincipal axes of stress and/or strain incrementare prevented from rotating and those in whichthey are free to rotate. The former category oftest includes the oedometer, the conventionaltriaxial tests, most versions of laboratory planestrain apparatus and the triaxial apparatus, andthe pressuremeter. The latter category includesthe direct simple shear test and most forms of insitu test, i.e. the vane shear test, the plate bear-ing test and the cone penetration test.

Since most in situ tests, allowing freedom ofthe principal axes to rotate, are correlated withlaboratory tests in which the principal axes areconstrained, it is important to assess the differ-ences in behaviour of soil under the two cir-cumstances. Additionally in all real engineeringsituations individual soil elements experiencestress and strain increments such that the princi-pal axes rotate, whereas nearly all mathematicalmodels of soil behaviour are based on data fromsingle-element tests in which no rotation is pre-sumed.Consequently a detailed study has been madeof the results of undrained direct simple shear

tests, and it is presented in this section. Thisstudy takes further the work of Randolph &Wroth (1981) concerned with the axial capacityof driven piles in clay and is based on thecomprehensive series of tests reported by Ladd& Edgers (1972), which have already been re-ferred to.

The tests were carried out in the Geonorapparatus in which cylindrical samples of soil arecontained in a rubber membrane reinforced in-ternally by a fine helical wire and subjected tosimple shear. Undrained tests are carried outas drained tests in which the total volume of thesample is held constant by suitable adjustmentof the vertical (effective) stress.Undrai ned t ests on normall y consoli datedspecimens

The results of a test on a specimen of nor-mally consolidated resedimented Boston blueclay are shown in the upper half of Fig. 22 inwhich the observed shear stress ri, (applied tothe horizontal top surface of the specimen) isplotted against the vertical effective stress oV.Both stress variables have been made dimen-sionless by dividing by the original consolidationpressure oVc. The individual points have beenplotted directly from the tabulated data in theMassachusetts Institute of Technology report.

The effective stress path ABC traced by thesepoints has two marked features. The first is thevery substantial reduction in effective verticalstress that occurs during the test, which wouldrequire very large positive excess pore pressuresto be generated in an equivalent undrained test.The second is that the maximum value of ob-served shear stress r,,, occurs at point B (after ashear strain of about 5% (see Fig. 25, later)),whereas the maximum stress ratio on the hori-zontal plane of the specimen occurs at point C(after 30% shear strain).

This raises the question whether failure ofthe specimen should be identified with state B orstate C. It will be shown later that state B doesnot represent the maximum shear stress on anyplane within the specimen; other planes canexperience greater values of shear stress at otherstages in the test. It is suggested therefore thatfailure in a frictional material should be as-sociated with the conditions of maximum stressratio.

The path ABC-and all other such paths ontests of seven normally consolidated clays re-ported by Ladd & Edgers (1972)-is nearlyelliptical in shape. Without attempting a physicalexplanation of why this might be so, it is worthexploring the consequences of assuming that allsuch paths can be closely matched by ellipses.


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474 I NTERP RETA T ION OF IN S I TU SOI L TESTS

(1 -sin 6)pG+Fig. 22. Undrained simple shear teat on normally consoli-dated Boston blue clay (data f r om test204, Ladd 8z Edgers,1972)

In the lower half of Fig. 22 a particular ellipseSTUVW has been drawn which passes throughthe initial point S and touches the failure en-velope OU at point U. To match the experimen-tal data as closely as possible the ellipse hasbeen drawn to the same scale and the failureenvelope has been chosen with the observedvalue of C$= c#J,,= 32.4. In theory there is aninfinite family of ellipses which satisfy thesecriteria each having a different centre and adifferent oblateness represented by the relativelength of the minor axis TW.

The particular ellipse chosen for simulation isgiven by the expression

4r2+ (a - h) = (a + h) sin &, (43)where for convenience 7 = ri,/uVvc, = a,/~,,and h = (1 -sin c#+)/( + sin &,.). This ellipse hassome special properties, which are discussedlater, and the abscissae of the main points inexplicit form are included in Fig. 22. In particu-lar the maximum value of shear stress, whichoccurs at point T, is given by the equation

7,,, 1 -sin 4,-= l+sin& tan 4,. (44)a,The observations made in a direct simple

shear test in a Geonor apparatus only provideinformation about the effective stresses on oneplane-the horizontal plane-in a supposedlyhomogeneous soil specimen. It is not possible todeduce the complete state of stress within thespecimen and to draw the relevant Mohr circle.Attempts have been made to meet this defi-ciency, notably by Soydemir (1976), by using thehelical reinforcement encased in the membraneas a transducer for meas

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