Geotechnical Evaluation Replacement Namaqua Bridge Over the … · 2018-11-28 · Typical Heel Drain Detail ... abutment and/or wing walls. Additionally 300 to 400 feet of Namaqua

Geotechnical Evaluation

Replacement Namaqua Bridge

Over the Big Thompson River

Loveland, Colorado

Revised 4

Prepared for:

Mr. Mike Oberlander

Interwest Consulting Group

1218 West Ash, Suite A

Windsor, Colorado 80550

Job Number: 16-3048 January 4, 2017

Namaqua Bridge

TABLE OF CONTENTS

Page

Purpose and Scope of Study ...................................................................................... 1

Proposed Construction ................................................................................................ 1

Site Conditions ............................................................................................................ 2

Geologic Setting ........................................................................................................... 3

Subsurface Exploration ............................................................................................... 4

Laboratory Testing ...................................................................................................... 4

Subsurface Conditions ................................................................................................ 5

Seismic Design Parameters ......................................................................................... 6

Geotechnical Consideration for Design .......................................................................... 7

Deep Foundations ........................................................................................................ 8

Shallow Foundations .................................................................................................. 17

Abutment and Wing Walls ............................................................................................ 19

Abutment Approaches .................................................................................................. 21

Lateral Loads ............................................................................................................... 21

Water-Soluble Sulfates ................................................................................................ 25

Soil Corrosivity ............................................................................................................. 27

Project Earthwork ....................................................................................................... 28

Excavation Considerations .......................................................................................... 32

Pavement Sections ....................................................................................................... 34

Closure ........................................................................................................................ 40

Locations of Test Holes ..................................................................................... Figure 1

Logs of the Test Holes ............................................................................... Figures 2 & 3

Legend and Notes ................................................................................................ Figure 4

Gradation Plots .......................................................................................... Figures 5 – 7

Direct Shear Data ................................................................................................ Figure 8

Seismic Response Spectrum ............................................................................. Figure 9

Capacity Reduction Factor Plots ........................................................... Figures 10 & 11

Typical Heel Drain Detail ................................................................................... Figure 12

Summary of Laboratory Test Results ........................................................ Tables 1 & 2

Core Hole Logs ............................................................................................. Appendix A

Pavement Section Calculations .................................................................... Appendix B

Namaqua Avenue Bridge Over the Big Thompson River

Loveland, Colorado Revised 4

Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 1

PURPOSED AND SCOPE OF STUDY

This report presents the results of a geotechnical evaluation performed by GROUND

Engineering Consultants, Inc. (GROUND), in support of design and construction of a

replacement bridge to carry Namaqua Avenue over the Big Thompson River near

Loveland, Colorado. Our study was conducted in general accordance with GROUND’s

Proposal No. 1511-2098 and associated contract agreement between Interwest

Consulting Group and GROUND, dated August 2, 2016.

A field exploration program was conducted to obtain information on subsurface

conditions. Material samples obtained during the subsurface exploration were tested in

the laboratory to provide data on the classification and engineering characteristics of the

on-site soils. The results of the field and laboratory studies are presented herein.

This report has been prepared to summarize the data obtained and to present our

conclusions based on the proposed construction and the subsurface conditions

encountered. Geotechnical design parameters and a discussion of geotechnical

engineering considerations related to the construction of the bridge also are included.

PROPOSED CONSTRUCTION

We understand that present plans call for either a 155-foot long, single-span bridge

structure, or a longer 3-span bridge structure that will carry Namaqua Avenue over the

Big Thompson River. We anticipate that that the bridge abutments likely will include

abutment and/or wing walls. Additionally 300 to 400 feet of Namaqua Avenue south of

the bridge and about 1,500 feet of Namaqua Avenue north of the bridge will be

reconstructed and widened to include sidewalks and bicycle trails. We also understand

that foundation design for the bridge will be following the current edition of AASHTO of

Load Resistance Factor Design (LRFD) methodology and that pavement design will be

developed using AASHTO (DARWin®) and Larimer County methodologies. If the use of

the current CDOT ME methodology is required, GROUND should be notified so that

appropriate revision can be made.

If the proposed construction differs significantly from that described above, GROUND

should be notified to re-evaluate our conclusions.




SITE CONDITIONS

The proposed bridge will span the Big Thompson River approximately 0.3 miles north of

Crestone Drive and approximately 0.5 miles south of US Highway 34. The surrounding

land was largely in use as farmland with associated single-family residences and farming

buildings (barns, sheds, etc.) or in use as sand and gravel pits for a concrete and

aggregate company. Namaqua Park was also observed to the southeast of the existing

bridge. Additionally, single-family residences and an elementary school were observed

in the greater project area.

The existing Namaqua Avenue was a 2-lane, asphalt-paved road and was aligned,

generally, north to south. Minor to moderate distress was observed in the existing

asphalt pavements (limited alligator cracking, local rutting, and minor transverse

cracking). The road grade elevation relative to the surrounding land near the crossing

varied locally and the road grade was raised between approximately 2 to 10 feet. A

retaining structure was observed along the northwest embankment of the road near the

existing bridge.

The Big Thompson River flowed to the east. The stream banks were relatively gentle on

the west (upstream) side of the bridge and moderately to relatively steep slopes on the

east (downstream) side of the bridge. These slopes displayed between approximately 3

feet (upstream side) and approximately 10 feet (downstream side) of relief. The

surrounding topography was generally rolling and sloped gently to the east near the

project site. Grasses and other relatively small native vegetation along with large,

mature trees were observed growing on the road shoulder and in the greater project

area.

Rip rap up to at least 2½ feet in dimension were noted along the existing bridge

abutments and bent, and below the flow-control structure on the downstream side of the

bridge. Rocks up to about 8 inches in dimension were noted among the native deposits.




GEOLOGIC SETTING

Published geologic maps, such as Colton (1978)1 depict the site as underlain by the

upper Holocene Post-Piney Creek Alluvium (Qpp) and the Pleistocene Broadway

Alluvium. In the project area, alluvium typically consists of sands, gravels, and boulders,

with silt and clay deposits locally. These surficial deposits are mapped as underlain by

the upper Cretaceous Lower Shale Member of the Pierre Shale Formation (Kpl). In the

project area, the Lower Shale Member of the Pierre Shale consists of the Mitten Black

Shale member, Sharon Springs Member, and Gammon Ferruginous which are largely

made up of dark olive-gray bentonitic shale. The shales typically are moderately to

highly expansive and are often very hard, well-cemented, and difficult to excavate.

A portion of the Colton (1978)1 geologic map is reproduced below.

1 Colton, R.B., 1978, Geologic map of the Boulder-Fort Collins-Greeley area, Front Range Urban Corridor,

Colorado: U.S. Geological Survey, Miscellaneous Investigations Series Map I-855-G, scale 1:100,000.

Approximate Project Alignment




SUBSURFACE EXPLORATION

Subsurface exploration for the project was conducted in August 2016. Two (2) test

holes were drilled at locations on either side of the bridge with a conventional, truck-

mounted, drilling rig to evaluate the subsurface conditions as well as to retrieve samples

for laboratory testing and analysis. The test holes were advanced to depths of about 41

to 50 feet below existing grades. A GROUND engineer directed the subsurface

exploration, logged the test holes in the field, and prepared the samples for transport to

our laboratory.

Samples of the subsurface materials were retrieved with a 2-inch I.D. California liner

sampler and a 1⅜-inch I.D. Standard Penetration Test sampler. The samplers were

driven into the substrata with blows from a 140-pound hammer falling 30 inches, in

general accordance with (in the case of the 1⅜-inch sampler) the Standard Penetration

Test described by ASTM Method D1586. Penetration resistance values, when properly

evaluated, indicate the relative density or consistency of soils. Depths at which the

samples were obtained and associated penetration resistance values are shown on the

test hole logs.

The approximate locations of the test holes relative to the existing Namaqua Avenue

bridge are shown in Figure 1. Logs of the test holes are presented in Figures 2 and 3,

and explanatory notes and a legend for the logs are provided in Figure 4. Logs of the

core runs are presented in Appendix A.

LABORATORY TESTING

Samples retrieved from our test holes were examined and visually classified in the

laboratory by the project engineer. Laboratory testing of soil samples included standard

property tests, such as natural moisture contents, dry unit weights, grain size analyses,

and Atterberg limits. Unconfined compressive strength, direct shear, water-soluble

sulfate content, swell-consolidation and a suite of corrosivity tests were completed on

selected samples, as well. Resilient modulus testing also was performed on a

composite sample obtained from the pavement test holes. Laboratory tests were

performed in general accordance with applicable ASTM and AASHTO protocols.

Results of the laboratory testing program are summarized in Tables 1 and 2 and on the




test hole logs. Gradation plots are provided in Figures 5 through 7 and may be used for

scour analysis. Direct Shear data is presented in Figure 8.

SUBSURFACE CONDITIONS

The test holes penetrated approximately 6 to 9 inches of asphalt before penetrating

sand, silt, and clay fill materials to depths between 4 and 11 feet below existing grade.

Additionally, a base course like material, approximately 12 inches in thickness was

encountered underlying the asphalt in Test Holes 1, 2, and P-1. Native sands and

gravels where encountered underling the fill materials that extended to depths of about

22 to 26 feet. These soils were underlain by clay shale bedrock which was encountered

at depths between 31 and 33 feet.

We interpret the native sands and gravels to be alluvial (stream-laid) deposits and the

bedrock clay shales to be Pierre Shale materials.

Fill soils were recognized in the test holes, and likely are present elsewhere near the

bridge site. Delineation of the complete lateral and vertical extents of fills at the site, or

their compositions, was beyond our present scope of services. If fill soil volumes and

compositions at the site are of significance, they should be evaluated using test pits.

Additionally, given the coarse nature of the materials encountered, it is difficult to readily

differentiate between coarse fill materials and base course materials. Section of base

course may be present beneath pavements where they were not indicated on the drill

logs.

Coarse gravel and boulders, as well as similarly sized fragments of debris are not

represented well in samples obtained from small diameter test holes. Therefore, such

coarse materials may be present, even where not included in the descriptions herein.

Given the general proximity of the site to the mountain front, the presence of boulders in

the subsurface in the deposits along river must be considered likely.

Fill fine to coarse sands and silts with local clays. They were dry to moist, non- to

slightly plastic, soft to very stiff or loose to dense, and brown dark brown to red brown in

color.

Sands and Gravels ranged from clean to clayey or silty, fine to coarse sands to that

included local boulders, and clay and silt beds. They were slightly moist to wet, non- to




slightly plastic, medium dense to very dense, and light brown to brown to gray brown in

color.

Clay Shale consisted of interbedded, non-durable, silt and clay shales. They were

slightly moist to moist, slightly to moderately plastic, very hard and dark gray to black in

color.

Groundwater was encountered in the test holes at depths of about 12 to 20 feet below

existing grade at the time of drilling. The test holes were immediately backfilled after

drilling operations due to safety concerns. Groundwater levels can be expected to

fluctuate, however, in response to annual and longer-term cycles of precipitation,

irrigation, surface drainage, land use, and the development of transient, perched water

conditions. We anticipate that water levels at the bridge abutments will tend to correlate

to water levels in the Big Thomson River, however.

SEISMIC DESIGN PARAMETERS

Based on extrapolation of available data to depth and our experience in the project area,

we consider the site likely to meet the criteria for a Seismic Site Classification of D

according to the 2014 AASHTO classification (Table 3.10.3.1-1). To evaluate the

Seismic Site Classification quantitatively, seismic shear wave velocity testing and/or

exploration to depths of at least 100 feet should be performed. We consider the

likelihood of achieving a Site Classification of C at the subject location to be moderate to

high. Seismic parameters that are applicable to this bridge site based on a Site

Classification of D are listed below:

Peak Ground Acceleration (PGA): 0.060 g

Short Period Spectral Acceleration (SS): 0.125 g

Long Period Spectral Acceleration (Sl): 0.032 g

FPGA: 1.6

Fa: 1.6

FV: 2.4

SDS: 0.200 g

SDl: 0.077 g

AASHTO Seismic Category: 1




A seismic response spectrum for the site, based on these parameters is provided in

Figure 9.

GEOTECHNICAL CONSIDERATIONS FOR DESIGN

Typically in the project area, heavy structures such as bridges are supported on deep

foundations bearing in the bedrock underlying the site. Pierre Shale Formation clay and

shales were encountered in foundation test holes at depths of about 22 to 26 feet

beneath the overlying alluvial and fill soils. Therefore, site appears suitable for

supporting the new bridge on driven pile foundations driven to refusal in the relatively un-

weathered bedrock. Anticipated pile tip depths are discussed below in the Deep

Foundations section of this report. Drilled piers also are feasible, but likely will be more

difficult to install.

We estimate that likely, post-construction movements of properly installed deep

foundations at the site will be about ½ inch.

Additionally, a Geosynthetic Reinforced Soil – Integrated Bridge System (GRS-IBS) may

be a cost effective alternative to driven piles. A GRS system is created by combining

geosynthetic reinforcement and a high quality granular fill to form a composite material

that is capable of carrying high bridge loads. This composite material is used in

conjunction with a GRS abutment and integrated bridge approaches to create the GRS

system. Likely post-construction movements for these systems are typically higher than

deep foundation systems, but a GRS-IBS may be easier and more cost effective to

construct. If GRS-IBS is selected, GROUND can provide design services for the

abutment and wing walls.

To limit differential movements relative to the bridge structure, abutment and wing walls

should be founded in a similar manner to the bridge. However, we understand that

shallow foundations are being considered for the abutment and wing walls.

Geotechnical parameters for design of shallow foundations for abutment and wing walls

are provided in the Shallow Foundations section of this report. We estimate post-

construction movements will be on the order of 1 inch for abutment and wing walls

supported on properly constructed shallow foundations.

Regardless of the selected foundation type, bridge elements should be designed to

account for the design depths of scour and associated loss of support.




DEEP FOUNDATIONS

Both driven steel H-piles and drilled piers, bearing in the bedrock underlying the

alignment appear suitable, in principle, to support the proposed bridge. We anticipate

that drilled pier may entail significantly greater difficulties to install. Geotechnical

parameters for both types of foundations are provided below.

The geotechnical parameters provided in this report are discussed in terms of the ‘North

Abutment’ and ‘South Abutment.’ If the 3-span design is selected, the parameters for

the North Abutment also are applicable to the northern bent and the parameters for the

Sorth Abutment also are applicable to the southern bent.

Note that the pile and drilled pier depths indicated herein refer to depths below existing

grades at the test hole locations at the time of drilling. The contractor should make

allowance for any grade changes between that time and the time of construction.

Lowering grades, however, may not be sufficient to result in shortening of piles,

however. We have assumed that pile caps / top-of-pile will be at depth of approximately

12 below existing grades.

Driven Pile Foundations Geotechnical criteria based on the AASHTO LRFD

methodology are provided below for design of driven, steel, H-pile foundations.

Driven piles should be installed per Section 502 of the current edition/revision of the

CDOT Standard Specifications for Road and Bridge Construction. The installation

considerations should be taken into account when preparing project documents.

Geotechnical Parameters for H-Pile Foundation Design

1) Driven steel H-piles to support the bridge and walls should be Grade 50, high

strength steel (ASTM A572) with a minimum yield strength of 50 ksi.

2) The piles should consist of heavy steel H-sections. We anticipate that HP 12x53,

HP 12x74, HP 14x73, or HP 14x89 sections will be used. Selection of pile size

should consider anticipated pile depths.

3) The piles should be driven to practical refusal, bearing in relatively un-weathered

bedrock underlying the site. The depths to which we anticipate that piles will be

driven are tabulated below:




ANTICIPATED PILE TIP ELEVATIONS

Location Test

Holes

Approximate

Top-of-Bedrock Depth

Anticipated Pile Tip

Depth for Moderately Loaded Piles

Anticipated Pile Tip

Depth for Heavily Loaded Piles

(feet below existing grade*)



Southern Abutment TH 3 & 4 23 ± 28 to 34 32 to 36

Northern Abutment TH 1 & 2 26 ± 31 to 37 35 to 39

*existing grade at time of drilling operations

Because of the potential variability depth to top-of-bedrock, the contractor should

be prepared to advance the piles to elevations at least 5 feet lower than the

depths tabulated above. A test pile program with dynamic pile testing (PDA)

would be beneficial for refining anticipated driving depths/elevations. At least one

test pile should be driven at each abutment.

A minimum pile penetration may need to be specified to meet the requirements

for uplift resistance, lateral resistance, and/or the depth to resist scour.

The structural engineer and contractor also should anticipate refusal locally

during pile installation. The contractor should be prepared to install additional

piles offset from refusal locations, as necessary.

4) Drivability analysis using a wave equation analysis should be performed to

establish preliminary installation criteria for driven piles.

5) Based on the structural strength limit (AASHTO Article 6.5.4) and drivability

analysis (AASHTO Article 10.7.8) for H-piles, the maximum driving stress should

not exceed the nominal structural strength of 36 ksi.

6) In order to allow pile dynamic testing (See ‘Installation Considerations,’ below.)

welded pile splices should be specified, and compression splices not allowed.

7) Geotechnical parameters for design of piles to resist lateral loads are provided in

the Lateral Loads section of this report.




8) Lateral resistance to horizontal forces also can be resisted by battered piles. It is

normal to assume a battered pile can resist the same axial load as a vertical pile

of the same type and size driven to the same depth. The vertical and horizontal

components of the load will depend on the batter inclinations.

Pile batters should not exceed 1:4 (horizontal : vertical).

9) For static analysis, based on the available geotechnical data and AASHTO Table

10.5.5.2.3-1, the nominal resistance values for axial compression and resistance

factors (φstat) tabulated below for each abutment may be used for pile design.

* Side resistance should be neglected for all material above the depth of design scour.

A higher soil resistance factor may be used in the pile foundation design, if the

driving criteria are established by static and/or dynamic testing. Provided that the

dynamic testing with signal matching per CDOT Section 502.05 is performed on

at least 1 pile at each abutment, then the nominal resistance values and dynamic

soil resistance factors (φdyn) tabulated below for each abutment may be used for

pile design:

Material

Depth Range

(feet)

North South

Loading

Type

Resistance

Type

Nominal

Resistance

Resistance

Factor

(φstat)

Fill 3* – 11 3* – 11 Axial Side 0.40 ksf 0.45

Uplift Side 0.40 ksf 0.35

Sands & Gravels 11* – 15 11* – 15 Axial Side 1.7 ksf 0.45




Shale Bedrock

26+

23+ Axial

Side 10 ksf 0.35

Tip 91 ksf 0.40

Uplift Side 10 ksf 0.25




* Side resistance should be neglected for all material above the depth of design scour.

For H-piles, the soil contact area should be applied for the skin resistance and

the box / square perimeter area should be applied for the tip resistance.

10) Groups of closely spaced piles will require an appropriate reduction of the

resistance values tabulated above. Reduction of axial capacities can be avoided

by spacing piles apart a distance of at least 3 ‘diameters’ center to center.

Reduction factors for piles spaced more closely than 3 diameters center to center

can be obtained from Figure 10.

11) Reduction of lateral capacity generally can be avoided by spacing piers at least 3

diameters apart, center to center. However, linear arrays of piles loaded in line

with the pile centers should be spaced at least 8 ‘diameters’ apart to avoid

reduction in axial capacity. Reduction factors for piles spaced more closely than

8 diameters can be obtained from Figure 11.

Installation Considerations

12) We suggest that a test pile installation program be performed to define better the

driving conditions, appropriate pile hammer energy, installation depth, and

refusal conditions. Due to the overburden soils being relatively thick at the site,

Material

Depth Range

(feet)

North South

Loading

Type

Resistance

Type

Nominal

Resistance

Resistance

Factor

(φdyn)

Fill 3* – 11 3* – 11 Axial Side 0.40 ksf 0.65






Shale Bedrock 26+ 23+ Axial Side 10 ksf 0.65

Tip 91 ksf 0.65





the wave equation analysis list in Item 4, above, may not always be able to

model the refusal conditions properly.

13) It may be beneficial to reinforce the piles with a commercial, heavy duty, pile tip.

14) The pile-driving hammer should develop a minimum of 26,000 foot-pounds of

energy per blow for HP 12x53 piles, 42,000 foot-pounds for HP 12x74 piles,

40,000 foot-pounds for HP 14x73 piles, and 52,000 foot-pounds for HP 14x89

piles. However, the hammer generally should not develop more than 2,500 foot-

pounds per square inch of cross sectional area (e.g., 38,750 foot-pounds for HP

12x53 piles) unless it is demonstrated by wave equation analysis that the piles

can be installed safely and efficiently.

15) After the actual pile type and proposed hammer have been selected, a

geotechnical engineer should be retained to perform a Wave Equation Analysis

to determine if the driving hammer is sized adequately for the type of pile

selected and the soils and bedrock materials into which the piles will be driven, if

a test pile program will not be performed.

16) Consideration also should be given to re-striking piles to evaluate their capacity

at least 24 hours after initial driving.

17) Pile dynamic testing should be performed using a pile driving analyzer (PDA) and

signal matching at the start of pile installation to:

Assess the condition of the pile,

Evaluate the efficiency of the hammer,

Measure the driving stress in the pile,

Determine the static capacity of the pile, and

Establish the pile driving criteria for required static capacity, or sands and gravels / bedrock penetration.

18) It may be beneficial also to perform lateral loading testing at the start of pile

installation.




19) Where a pile cannot be advanced to at least the approximate, anticipated tip

elevation, it should be evaluated with regard to its capacity by the geotechnical

engineer and the structural engineer.

Drilled Pier Foundations Geotechnical criteria based on the AASHTO LRFD

methodology are provided below for design of straight-shaft, drilled pier foundations.

Geotechnical Parameters for Drilled Pier Design

1) Drilled piers should bear in relatively un-weathered bedrock underlying the site.

For design purposes, ‘relatively un-weathered’ bedrock may be taken to be at

and below depths of 23 feet below existing grades at the north abutment and at

and below 26 feet at the south abutment. For bidding purposes this may vary

somewhat at each abutment.

2) Drilled piers should be at least 24 inches in diameter and generally should be

designed for a maximum length to diameter ratio of 30:1. The actual pier

diameters and length to diameter ratios should be determined by the structural

engineer, however.

3) Drilled piers should have a minimum length of 29 feet at the north abutment and

32 feet at the south abutment. The actual pier lengths should be determined by

the structural engineer based on loading, etc., with further increases in length

possibly required by the conditions encountered during installation at each pier

location.

4) Piers also should penetrate at least 6 feet into relatively un-weathered bedrock

or 3 pier diameters, whichever is greater.

Based on a the minimum length and bedrock penetration, and taking top of

competent bedrock to be 23 feet (north abutment) or 26 feet (south abutment)

below grade, pier lengths of 31 to 34 feet at the north abutment or 34 to 37 at the

south abutment are anticipated to meet the geotechnical criteria. Actual pier

lengths commonly will be greater due to structural considerations, conditions in

the pier holes, etc.




5) Drilled piers may be designed using the nominal resistance values and

resistance factors tabulated below for each abutment.

Material

Depth Range

(feet)

North South

Loading

Type

Resistance

Type

Nominal

Resistance

Resistance

Factor

Fill 5* – 11 5* – 11 Axial Side 0.35 ksf ** 0.55

Uplift Side 0.35 ksf ** 0.45

Sands & Gravels 11* – 15 11* – 15 Axial Side 0.95 ksf ** 0.55


Sands & Gravels 15* – 26 15* – 23 Axial Side 1.15 ksf ** 0.55


Shale Bedrock

26+

23+ Axial

Side 11 ksf 0.45

Tip 120 ksf 0.50


* Side resistance should be neglected for all material above the depth of design scour. In addition, AASHTO indicates that the uppermost 5 feet should be neglected for calculating side resistance.

** Assumed to be cased during installation.

6) Estimated settlement of properly constructed drilled piers will be low, on the order

of ½ inch, to mobilize side resistance. Settlement of groups of piers spaced

more closely than 3 diameters, center to center, may be larger and should be

studied under an individual basis.

7) Piers should be reinforced adequately, as determined by the structural engineer.

8) Geotechnical parameters for resisting lateral loading of piers are provided in the

Lateral Loads section of this report.

9) Rock penetration in pier holes should be roughened artificially to assist the

development of peripheral shear between the pier and bedrock. Artificially

roughening of pier holes should consist of installing shear rings 3 inches high

and 2 inches deep in the portion of each pier penetrating bedrock below depths

of 23 feet at the north abutment or 26 feet at the south abutment. The shear

rings should be installed on 18-inch centers.




The specifications should allow a geotechnical engineer to waive the requirement

for shear rings depending on the conditions actually encountered in individual

pier holes, however.

10) Groups of closely spaced drilled piers will require an appropriate reduction of the

resistance values tabulated above. Reduction of resistance values can be

avoided by spacing piers at least 3 diameters apart, center to center. Reduction

factors for the resistance values for piers spaced more closely than 3 diameters

center to center can be obtained from Figure 10.

11) Reduction of lateral capacity generally can be avoided by spacing piers at least 3

diameters apart, center to center. However, linear arrays of drilled piers loaded

in line with the array (parallel to the line connecting the pier centers) must be

spaced at least 8 diameters apart, center to center, to avoid reductions in lateral

capacity. Reduction factors for lateral capacities for closely spaced drilled piers

can be obtained from Figure 11.

Drilled Pier Construction Considerations

12) We anticipate that due to the relatively permeable nature of the fill and native

sands overlying the bedrock at the site, and the presence of coarse gravel and

boulders in these materials associated with the Big Thompson River, advancing

the drilled pier holes to the design depths or more will be relatively difficult.

13) The top of competent bedrock should be determined in the field at each pier

location. For the purpose of assessing pier length and bedrock penetration, the

top of competent bedrock should not be counted as shallower than the design

depths discussed above, even where encountered at a shallower depth.

14) Lenses or beds of relatively soft bedrock not suitable for foundation support may

be encountered within the competent bedrock section. Construction observation

of such materials penetrated by the pier excavations will require deepening of

individual piers.

15) The bedrock beneath the bridge alignment was very hard and resistant.

Penetration-resistance values of 50 blows for 2 inches were typical. The pier-

drilling contractor should mobilize equipment of sufficient size and operating




capability to advance the pier holes through deposits including coarse gravel and

boulders, and achieve the design lengths and bedrock penetration. The

contractor should be prepared to core highly resistant bedrock materials.

If refusal is encountered in these materials, a geotechnical engineer should be

retained to evaluate the conditions to establish whether true refusal has been

met with adequate drilling equipment.

16) Groundwater was encountered during subsurface exploration at depths between

about 12 and 20 feet. Water levels presumably will vary with stream stage.

Groundwater will be encountered during drilled pier installation. Casing likely will

be required in the pier holes to reduce water infiltration. In the event that casing

is seated into the bedrock, the minimum bedrock penetration should be taken

from the bottom of the casing.

Seating of the casing in the upper layers of the bedrock may not create positive

cutoff of water infiltration. The contractor should be prepared to address this

condition.

17) In no case should concrete be placed in more than 3 inches of water, unless

placed through an approved tremie method. The proposed tremie method be

discussed during the pre-construction meeting by the Project Team.

18) Where groundwater and unconsolidated soils are encountered, the installation

procedure of drilled piers can be a concern. Commonly in these conditions, the

drilling contractor utilizes casing and slurry during excavation of the pier holes,

which may adversely affect the axial and/or lateral capacities of the completed

piers. During casing withdrawal, the concrete have sufficient slump and must be

maintained with sufficient head above groundwater levels to displace the water or

slurry fully to prevent the creation of voids in the pier.

Because of these considerations, the drilling contractor should submit a written

procedure addressing the use of casing, slurry, and concrete placement prior to

commencement of drilled pier installation.

19) Pier holes should be properly cleaned prior to placement of concrete.




20) Concrete utilized in the piers should be a fluid mix with sufficient slump so that it

will fill the void between reinforcing steel and the pier hole wall, and help exclude

soil, water and slurry from entering the concrete. Concrete should have a

minimum slump in the range of 5 to 7 inches.

21) Concrete should be placed by an approved tremie method or other method to

reduce mix segregation.

22) Concrete should be placed in a pier on the same day that it is drilled. Failure to

place concrete the day of drilling normally will result in a requirement for

lengthening the pier. The presence of groundwater or caving soils may require

that concrete be placed immediately after the pier hole drilling is completed.

23) The contractor should take care to prevent enlargement of the excavation at the

tops of piers, which could result in “mushrooming” of the pier top. Mushrooming

of pier tops can increase uplift pressures on the piers.

24) Sonic integrity testing (sonic echo or cross-hole sonic) should be performed for

an appropriate percentage of the drilled piers to assess the effectiveness of the

pier construction methods for installing the piers in accordance with project plans

and specifications. Testing 10 percent of drilled piers is common, with testing of

additional piers if voids or other flaws are detected.

Additional information on sonic integrity testing can be provided upon request.

SHALLOW FOUNDATIONS

The geotechnical criteria below should be observed for spread footing foundation

systems for the project retaining walls.

1) The footings should bear on dense, undisturbed, native sands and gravels, or on

‘dental’ concrete. Footing bearing elevations may need to be lowered to

account for scour or because of the conditions exposed at the actual abutment

locations.

Fine-grained, soft, loose, or otherwise deleterious materials exposed at footing

bearing elevation, e.g., a silt pocket, should be removed entirely, and replaced




with ‘dental’ concrete. Likewise, voids created by removal of boulders should be

replaced with ‘dental’ concrete.

A geotechnical engineer should be retained to observe the surfaces on which the

footings will bear. The exposed surfaces should be approved prior to placement

of reinforcing steel or footing concrete.

2) Based on the conditions encountered at the test holes, footings bearing on

dense, undisturbed, native sands and gravels, or on ‘dental’ concrete, may be

designed for the nominal bearing pressure and resistance factor tabulated below.

Nominal Bearing Pressure Resistance Factor (φb)

10,500 psf 0.45

Compression of the bearing soils for the provided nominal bearing pressure is

estimated to be 1 inch, based on an assumption of drained foundation conditions.

3) In order to reduce differential settlements between footings along continuous

footings, footing loads should be as uniform as possible. Differentially loaded

footings will settle differentially. Similarly, differential fill thicknesses beneath

footings will result in increased differential settlements.

4) Spread footings should have a minimum lateral dimension of 48 inches. Actual

footing dimensions should be determined by the structural engineer.

3) Footings should bear at an elevation below the design scour elevation or

required frost protection depth. Footings should bear at least 3 feet below lowest

adjacent grade for protection against frost; bearing at a greater depth may be

necessary for scour protection.

4) Continuous foundation walls should be reinforced as determined by the structural

engineer.

5) Geotechnical parameters for lateral resistance to foundation loads are provided

in the Lateral Earth Pressures section of this report.




Shallow Foundation Construction

6) The contractor should take adequate care when making excavations not to

compromise the bearing or lateral support for nearby improvements.

7) Footing excavation bottoms may expose fine grained soils, loose, soft, or

otherwise deleterious materials. Firm materials may become disturbed by the

excavation process. All such unsuitable materials should be excavated and

replaced with ‘dental’ concrete.

8) Fill placed against the sides of the footings should be properly compacted in

accordance with the Project Earthwork section of this report.

ABUTMENT AND WING WALLS

To achieve similar performance, any abutment or wing walls for the bridge should be

supported on drive steel H-pile foundations in the same manner as the bridge, or

supported on the bridge foundations. Geotechnical parameters for driven pile and drilled

pier foundation systems are provided in the Deep Foundations section of this report.

Shallow foundations also appear feasible to support the abutment and wing walls.

Geotechnical parameters for these are provided in the Shallow Foundations section of

this report.

Lateral load parameters for design of the abutment and wing walls for the bridge are

provided in the Lateral Loads section of this report. Wall design should incorporate any

upward sloping backfills, live loads such as traffic, construction equipment, material

stockpiles, etc., and other surcharge pressures.

Wall Drainage Abutment and/or wing walls should be provided with drains at the heels

of the walls. In addition, damp-proofing should be applied to the backs of the walls and

Tencate MiraFi® G-Series backing (or comparable wall drain provisions such as the use

of granular wall backfill) should be placed on the backs of the walls, extending to within

about 2 feet of the top-of-wall. The wall drain system should be in hydraulic continuity

with the heel drain. Each heel drain system should be tested by the contractor after

installation and after placement and compaction of the overlying backfill to verify that the

system functions properly. A typical heel drain detail is provided in Figure 12.




Geotechnical Parameters for Wall Heel Drain Design Heel drains for the walls should be

designed in accordance with the parameters below. The actual underdrain layout,

outlets, and locations should be developed by wall designer.

1) A heel drain should consist of perforated, rigid, PVC collection pipe at least 3

inches in diameter, non-perforated, rigid, PVC discharge pipe at least 3 inches

in diameter, free-draining gravel, and filter fabric.

2) The free-draining gravel should contain 5 or less percent passing the No. 200

Sieve and 50 or more percent retained on the No. 4 Sieve, and have a

maximum particle size of 2 inches or meet the requirements of CDOT Class B

Filter Material. Each collection pipe should be surrounded on the sides and top

with 6 or more inches of free-draining gravel.

3) The gravel surrounding the collection pipe(s) should be wrapped with filter fabric

(MiraFi 140N® or the equivalent) to reduce the migration of fines into the drain

system.

4) The heel drain system should be designed to discharge at least 15 gallons per

minute of collected water.

5) The discharge point(s) for the collection pipe flow lines should be above the

design ditch stage. (Wall design should incorporate hydrostatic loads on the

portion of the wall below the ditch.)

The collection and discharge pipe for the heel drain system should be laid on a

slope sufficient for effective drainage, but a minimum of 1 percent.

6) ‘Clean-outs’ should be provided for each heel drain to facilitate maintenance of

the underdrains.

7) The discharge piping should be connected to one or more sumps from which

water can be removed by pumping, or to outlet(s) for gravity discharge.




ABUTMENT APPROACHES

We anticipate that several feet of new fill may be placed to construct the bridge

abutment approaches. Parameters and standards for fill placement and compaction, as

well as fill settlement estimates, are provided in the Project Earthwork section of this

report. Because the abutment fills necessarily are of differential thickness, settlements

generally will be differential with distance from the abutment.

Surface settlements also will depend on the depth and compressibility of the foundation

materials. Penetration resistance values at relatively shallow depths in the test holes

indicate that the fill and native overburden materials underlying the proposed abutment

approach areas at generally are stiff to very stiff or loose to medium dense. Therefore,

we anticipate additional settlement of approximately 1 inch in the new fill soils and the

native soils underlying the abutment approach fills. Differential settlements likely will be

proportional to fill depth.

Bridge and approach design should consider including “sleeper slabs” beneath the

roadway at both abutments to accommodate the transition from bridge structure to filled

ground. “Sleeper slabs” should not be installed until settlement of the approach fill

materials is substantially complete. GROUND estimates that the majority of settlement

will occur within 2 to 3 months after the completion of approach fill placement. Should

scheduling constraints necessitate earlier construction of settlement sensitive structures,

the approach fill(s) should be monitored weekly by surveying the top of the fill near the

abutment. When three successive measurements indicate elevation changes that are

less than ¼-inch, the majority of settlement may be considered to be complete.

LATERAL LOADS

Based on the data obtained for this study and our experience with similar sites and

conditions, lateral load analysis using the Terzaghi or Brown method may take the

values tabulated below for the modulus of horizontal subgrade reaction (Kh) to be

characteristic of the soils and bedrock underlying the site, based on a simplified soil /

bedrock profile. Resistance to lateral loads by deep foundations should be neglected in

the upper 3 feet of soils, whether fill or native.




HORIZONTAL MODULUS SUBGRADE REACTION (Kh) – TERZAGHI METHOD

Soil / Bedrock Material

Approximate Depth Range

North Abut.


South Abut.

Kh based on Foundation

Element Width / Diameter

12-inch 14-inch 24-inch 30-inch

Fill 3* – 11 feet 3* – 11 feet 54 tcf

(62 pci)

46 tcf

(53 pci)

27 tcf

(31 pci)

22 tcf

(25 pci)

Sands & Gravels

11* – 15 feet 11* - 15 feet 180 tcf

(208 pci)

154 tcf

(178 pci)

90 tcf

(104 pci)

72 tcf

(83 pci)

Sands & Gravels

15* – 26 feet 15* – 23 feet 300 tcf

(347 pci)

257 tcf

(297 pci)

150 tcf

(174 pci)

120 tcf

(139 pci)

Shale Bedrock

23+ feet 26+ feet 450 tcf

(520 pci)

386 tcf

(446 pci)

225 tcf

(260 pci)

180 tcf

(208 pci)

* Lateral resistance should be neglected Note that the Kh values tabulated above are dependent on deep foundation

for material above the depth of design element width or diameter. Values for other widths / diameters can be provided

scour. upon request.

If “L-Pile” or a similar computer program is used for lateral analysis of the piles, the

geotechnical parameters tabulated below may be used for input into that program, and

are for the same simplified soil / bedrock profile. These include, unit wet weights (γ'), angle of internal friction (), and cohesion (c) for the earth materials, as well as values

for strain at 50 percent of failure stress (50) and horizontal soil modulus (k). Again,

resistance to lateral loads should be neglected in the upper 3 feet of fill or native soils.




GEOTECHNICAL PARAMETERS FOR LATERAL LOAD ANALYSIS USING L-PILE

Soil / Bedrock Material


North Abut.


South Abut.

Parameter

Value

Fill

(model as Sand without Free Water) 3* – 11 feet 3* – 11 feet

γ' 115 pcf

30 degrees

k 0.0622 x 106 pcf

(36 pci)

Sands & Gravels

above the Water Table

(model as Sand without Free Water)

11* – 15 feet

11* – 15 feet

γ' 118 pcf

33 degrees

k 0.289 x 106 pcf

(167 pci)

Sands & Gravels

below the Water Table

(model as Sand with Free Water)

15* – 26 feet

15* – 23 feet

γ' 65 pcf

36 degrees

k 0.480 x 106 pcf

(278 pci)

Claystone Bedrock

(model as Stiff Clay w/o Free Water)

26+ feet

23+ feet

γ' 127 pcf

c 8,900 psf

50 0.004

* Lateral resistance should be neglected for material above the scour depth.

Shallow Foundations Resisting Lateral Loads Cast-in-place footings bearing on

dense, undisturbed, native sands and gravels, or on ‘dental’ concrete, and backfilled with

similar materials may be designed for the parameters tabulated below. The parameters

below also apply to abutment and wing walls. If select, granular imported soil is used as

wall backfill, it should meet the parameters for CDOT Class 1 Structure Backfill.

The lateral earth pressures tabulated below may be used for design of the shallow

foundations,




The upper 1 foot of embedment should be neglected for passive resistance, however.

Where this passive soil pressure is used to resist lateral loads, it should be understood

that significant lateral strains will be required to mobilize the full value indicated above,

likely 1 inch or more. A reduced passive pressure can be used for reduced anticipated

strains, however.

The values for site soils were based on a moist unit weight (γ') of 120 pcf and an angle

of internal friction () of 32 degrees. The values for CDOT Class 1 Structure Backfill

were based on a moist unit weight (γ') of 132 pcf and an angle of internal friction () of

34 degrees.

EQUIVALENT FLUID WEIGHTS (DRAINED CONDITION)

Backfill

Material

Condition

Active

At-Rest

Passive

Site Soils 38 pcf 57 pcf 390 pcf (to a maximum of 3,900 psf)

CDOT Class 1 Structure Backfill

38 pcf 59 pcf -

EQUIVALENT FLUID WEIGHTS (SUBMERGED CONDITION)

Backfill

Material

Condition

Active

At-Rest

Passive

Site Soils 81 pcf 90 pcf 250 pcf (to a maximum of 2,500 psf)

CDOT Class 1 Structure Backfill

83 pcf 94 pcf -

RESISTANCE FACTORS

Sliding Resistance Passive Resistance

Φτ tan δ Φep

0.80 0.625 0.50




Wall design should incorporate any upward sloping backfills, live loads such as

construction equipment, material stockpiles, etc., and other surcharge pressures. The

build-up of hydrostatic pressures behind a wall also will increase lateral earth pressures

on the walls.

If select, granular fill is placed as abutment or wing wall backfill, in order to realize the

(lower) values for that material, then the select granular fill should be placed behind the

wall to a minimum distance equal or greater than half of the wall height.

The criteria for CDOT Class 1 Structure Backfill are tabulated in the Project Earthwork

section of this report.

WATER-SOLUBLE SULFATES

The concentrations of water-soluble sulfates measured in samples of the site soils

ranged up to 0.46 percent by weight. (See Table 2.) Such concentrations of soluble

sulfates represent a severe environment for sulfate attack on concrete exposed to these

materials. Degrees of attack are based on the scale of 'negligible,' 'moderate,' 'severe'

and 'very severe' as described in the “Design and Control of Concrete Mixtures,”

published by the Portland Cement Association (PCA). The Colorado Department of

Transportation (CDOT) utilizes a corresponding scale with four classes of severity of

sulfate exposure (Class 0 to Class 3) as described in the table below.

REQUIREMENTS TO PROTECT AGAINST DAMAGE TO CONCRETE BY SULFATE ATTACK FROM EXTERNAL SOURCES OF SULFATE

Severity of Sulfate

Exposure

Water-Soluble Sulfate (SO4

=) In Dry Soil

(%)

Sulfate (SO4=)

In Water

(ppm)

Water / Cementitious Ratio

(maximum)

Cementitious Material

Requirements

Class 0 0.00 to 0.10 0 to 150 0.45 Class 0

Class 1 0.11 to 0.20 151 to 1500 0.45 Class 1

Class 2 0.21 to 2.00 1501 to 10,000 0.45 Class 2

Class 3 2.01 or greater 10,001 or greater 0.40 Class 3

Based on our test results and PCA and CDOT guidelines, sulfate-resistant cement

should be used in all concrete exposed to site soil and bedrock, conforming to one of the

following Class 2 requirements:




(1) ASTM C 150 Type V with a minimum of a 20 percent substitution of Class F fly

ash by weight

(2) ASTM C 150 Type II or III with a minimum of a 20 percent substitution of Class F

fly ash by weight. The Type II or III cement shall have no more than 0.040

percent expansion at 14 days when tested according ASTM C 452

(3) ASTM C 1157 Type HS; Class C fly ash shall not be substituted for cement.

(4) ASTM C 1157 Type MS plus Class F fly ash where the blend has less than 0.05

percent expansion at 6 months or 0.10 percent expansion at 12 months when

tested according to ASTM C 1012.

(5) A blend of Portland cement meeting ASTM C 150 Type II or III with a minimum of

20 percent Class F fly ash by weight, where the blend has less than 0.05 percent

expansion at 6 months or 0.10 percent expansion at 12 months when tested

according to ASTM C 1012.

(6) ASTM C 595 Type IP(HS); Class C fly ash shall not be substituted for cement.

When fly ash is used to enhance sulfate resistance, it shall be used in a proportion

greater than or equal to the proportion tested in accordance to ASTM C 1012, shall be

the same source, and it shall have a calcium oxide content no more than 2.0 percent

greater than the fly ash tested according to ASTM C 1012.

All concrete exposed to site soil and bedrock should have a minimum compressive

strength of 4,500 psi.

The contractor should be aware that certain concrete mix components affecting sulfate

resistance including, but not limited to, the cement, entrained air, and fly ash, can affect

workability, set time, and other characteristics during placement, finishing and curing.

The contractor should develop mix(es) for use in project concrete which are suitable with

regard to these construction factors, as well as sulfate resistance. A reduced, but still

significant, sulfate resistance may be acceptable to the owner, in exchange for desired

construction characteristics.




SOIL CORROSIVITY

Data were obtained to support an initial assessment of the potential for corrosion of

ferrous metals in contact with earth materials at the site, based on the conditions at the

time of GROUND’s evaluation. The test results are summarized in Table 2.

Soil Resistivity A sample of materials retrieved from the test hole was tested for

resistivity in the laboratory, at approximately in-place moisture contents. Measurement

of electrical resistivity indicated a value of approximately 953 ohm-centimeters in a

sample of site soils.

pH Where pH is less than 4.0, soil serves as an electrolyte; the pH range of about 6.5 to

7.5 indicates soil conditions that are optimum for sulfate reduction. In the pH range

above 8.5, soils are generally high in dissolved salts, yielding a low soil resistivity.2

Testing indicated a pH value of about 8.4.

Reduction-Oxidation testing indicated red-ox potentials of about -95 millivolts. Such

low potentials typically indicate a more corrosive environment.

Sulfide Reactivity testing indicated “positive” results in the site soils. The presence of

sulfides in the soils suggests a more corrosive environment.

Corrosivity Assessment The American Water Works Association (AWWA) has

developed a point system scale used to predict corrosivity. The scale is intended for

protection of ductile iron pipe but is valuable for project steel selection. When the scale

equals 10 points or higher, protective measures for ductile iron pipe are indicated. The

AWWA scale is presented below. The soil characteristics refer to the conditions at and

above pipe installation depth.

We anticipate that drainage at the site after construction will be effective. Nevertheless,

based on the values obtained for the soil parameters, the overburden soils and bedrock

appear to comprise a severely corrosive environment for ferrous metals (18½ points).

If additional information is needed regarding soil corrosivity, GROUND suggests

contacting the American Water Works Association or a Corrosion Engineer. It should be

2 American Water Works Association ANSI/AWWA C105/A21.5-05 Standard




noted, however, that changes to the site conditions during construction, such as the

import of other soils, might alter corrosion potentials significantly.

Table A.1 Soil-test Evaluation

Soil Characteristic / Value Points

Soil Resistivity

<1,500 ohm-cm ..........................................................................................… 10 1,500 to 1,800 ohm-cm ................................................................……......…. 8 1,800 to 2,100 ohm-cm .............................................................................…. 5 2,100 to 2,500 ohm-cm ...............................................................................… 2 2,500 to 3,000 ohm-cm .................................................................................. 1 > 3,000 ohm-cm ................................................................................… 0

pH

0 to 2.0 ............................................................................................................ 5 2.0 to 4.0 ......................................................................................................... 3 4.0 to 6.5 ......................................................................................................... 0 6.5 to 7.5 ......................................................................................................... 0 * 7.5 to 8.5 ......................................................................................................... 0 > 8.5 .......................................................................................................... 3

Redox Potential

< 0 (negative values) ....................................................................................... 5 0 to +50 mV ................................................................................................…. 4 +50 to +100 mV ............................................................................................… 3½ > +100 mV ............................................................................................... 0

Sulfide Presence

Positive ........................................................................................................…. 3½ Trace .............................................................................................................… 2 Negative .......................................................................................................…. 0

Moisture

Poor drainage, continuously wet ..................................................................…. 2 Fair drainage, generally moist ....................................................................… 1 Good drainage, generally dry ........................................................................ 0 * If sulfides are present and low or negative redox-potential results (< 50 mV) are obtained, add three points for this range.

PROJECT EARTHWORK

We anticipate that earthwork construction at the site will consist largely of excavation

and backfilling at the bridge abutments. The earthwork criteria and standards below are

directed toward that work and are based on our interpretation of the geotechnical

conditions encountered in the test holes. Where these standards differ from applicable

municipal or agency specifications, e.g., for compaction of utility trench backfill, the latter

should be considered to take precedence.




General Considerations Prior to earthwork construction, existing concrete, vegetation

and other deleterious materials should be removed and disposed of off-site. Relic

underground utilities should be abandoned in accordance with applicable regulations,

removed as necessary, and properly capped.

Topsoil present on-site should not be incorporated into ordinary fills. Instead, topsoil

should be stockpiled during initial grading operations for placement in areas to be

landscaped or for other approved uses.

Existing Fill Soils Fill soils were recognized at in the test holes and likely are present

elsewhere in the vicinity, as well. All existing fill soils on which project elements bear

should be excavated entirely and replaced by properly compacted fill.

Although the majority of existing fill soils appeared suitable for re-use as compacted fill, it

is possible that some existing fill materials when excavated may not be suitable for re-

use. Debris and other deleterious materials may be encountered. Excavated fill

materials should be evaluated and tested, as appropriate, with regard to re-use.

Use of Existing Native Soils The local native soils that are free of trash, organic

material, construction debris, and other deleterious materials, are suitable, in general, for

placement as compacted fill. Significant quantities of organic materials should not be

incorporated into project fills.

Fragments of rock and boulders, (as well as inert construction debris, e.g., concrete or

asphalt) up to 3 inches in maximum dimension may be included in project fills, in

general. Such materials should be evaluated on a case-by-case basis where identified

during earthwork.

Imported Fill Materials If it is necessary to import material to the site as common fill,

the imported soils should be free of organic material, and other deleterious materials.

Imported material should consist of soils that exhibit 40 percent or less passing the No.

200 Sieve and a plasticity index of 10 or less. Representative samples of the materials

proposed for import should be tested and approved prior to transport to the alignment.

Material to be imported to the site as select, granular fill should meet the criteria for

CDOT Class 1 Structure Backfill (tabulated below).




CDOT CLASS 1 STRUCTURE BACKFILL

Sieve Size or Parameter Acceptable Range

2-inch 100% passing

No. 4 30% to 100% passing



Liquid Limit < 35

Plasticity Index < 6

Soils proposed for import as select, granular fill should be tested and approved prior to

transport to the site.

Fill Platform Preparation Prior to filling, the top 12 inches of in-place materials on

which fill soils will be placed should be scarified, moisture conditioned, and properly

compacted in accordance with the criteria below to provide a uniform base for fill

placement.

If surfaces to receive fill expose loose, wet, soft, or otherwise deleterious material,

additional material should be excavated, or other measures taken to establish a firm

platform for filling. A surface to receive fill must be effectively stable prior to placement

of fill.

Wet, Soft, or Unstable Subgrades Where wet, soft, or unstable subgrades are

encountered, the contractor must establish a stable platform for fill placement and

achieving compaction in the overlying fill soils. Therefore, excavation of the unstable

soils and replacing them with relatively dry or granular material, possibly together with

the use of stabilization geo-textile or geo-grid, may be necessary to achieve stability.

Whereas the stabilization approach should be determined by the contractor, GROUND

offers the alternatives below for consideration. Proof-rolling can be beneficial for

identifying unstable areas.

Replacement of the existing subgrade soils with clean, coarse, aggregate (e.g.,

crushed rock or “pit run” materials) or road base. Excavation and replacement to a

depth of 1 to 2 feet commonly is sufficient, but greater depths may be necessary to

establish a stable surface.




On very weak subgrades, an 18- to 24-inch “pioneer” lift that is not well compacted

may be beneficial to stabilize the subgrade. Where this approach is employed,

however, additional settlements of up to ½ inch may result.

Where coarse, aggregate alone does not appear sufficient to provide stable

conditions, it can be beneficial to place a layer of stabilization geo-textile or geo-

grid (e.g., Tencate Mirafi® HP370 or RS 580i, or Tensar® BX 1100) at the base of

the aggregate section.

The stabilization geo-textile / geo-grid should be selected based on the aggregate

proposed for use. It should be placed and lapped in accordance with the

manufacturer’s recommendations.

Geo-textile or geo-grid products can be disturbed by the wheels or tracks of

construction vehicles; care be taken to maintain the effectiveness of the system.

Placement of a layer of aggregate over the geo-textile / geo-grid prior to allowing

vehicle traffic over it can be beneficial in this regard.

When a given remedial approach has been selected, GROUND suggests constructing a

test section to evaluate the effectiveness of the approach prior to use over a larger area.

General Considerations for Fill Placement Fill soils should be thoroughly mixed to

achieve a uniform moisture content, placed in uniform lifts not exceeding 8 inches in

loose thickness, and properly compacted. No fill materials should be placed, worked,

rolled while they are frozen, thawing, or during poor/inclement weather conditions.

Where soils supporting foundations or on which foundation will be placed are exposed to

freezing temperatures or repeated freeze – thaw cycling during construction – commonly

due to water ponding in foundation excavations – bearing capacity typically is reduced

and/or settlements increased due to the loss of density in the supporting soils. After

periods of freezing conditions, the contractor should re-work areas affected by the

formation of ice to re-establish adequate bearing support.

Care should be taken with regard to achieving and maintaining proper moisture contents

during placement and compaction. Materials that are not properly moisture conditioned

may exhibit significant pumping, rutting, and deflection at moisture contents near




optimum and above. The contractor should be prepared to handle soils of this type,

including the use of chemical stabilization, if necessary.

Compaction areas should be kept separate, and no lift should be covered by another

until relative compaction and moisture content within the indicated ranges are obtained.

Compaction Standards Soils that classify as A-1, A-2 or A-3 soils in accordance with

the AASHTO classification system (granular materials), including select, granular fill,

should be compacted to 95 percent or more of the maximum dry density at moisture

contents at or above 2 percent below of the optimum moisture content as

determined by AASHTO T 180, the ‘modified Proctor.’

Soils that classify as A-4, A-5, A-6 or A-7 should be compacted to at least 95 percent of

the maximum dry density at moisture contents from at or above 2 percent below the

optimum moisture content as determined by AASHTO T 99, the ‘standard Proctor.’

Settlements Settlements will occur in filled ground, typically on the order of 1 to 2

percent of the fill depth. For a 12-foot fill, for example, that corresponds to a total

settlement of about 2 inches. If fill placement is performed properly, in GROUND’s

experience the majority (on the order of 60 to 80 percent) of that settlement typically will

take place during earthwork construction, provided the contractor achieves the

compaction levels indicated herein. The remaining potential settlements likely will take

several months or longer to be realized, and may be exacerbated if these fills are

subjected to changes in moisture content.

Cut and Filled Slopes Permanent, graded slopes supported by local soils up to 15 feet

in height should be constructed no steeper than 3:1 (horizontal : vertical). Minor

raveling or surficial sloughing should be anticipated on slopes cut at this angle until

vegetation is well re-established. Surface drainage should be designed to direct water

away from slope faces into designed drainage pathways or structures.

EXCAVATION CONSIDERATIONS

Excavation Difficulty Test holes for the subsurface exploration were advanced to the

depths indicated on the test hole logs by means of conventional, truck-mounted,

geotechnical, drilling equipment. We anticipate no unusual excavation difficulties in

these materials, in general, for the proposed construction with conventional, heavy duty,




excavating equipment. Excavations may encounter coarse gravel or boulders, however.

Fill soils may include coarse construction debris, as well.

Temporary Excavations and Personnel Safety Excavations in which personnel will

be working must comply with all applicable OSHA Standards and Regulations,

particularly CFR 29 Part 1926, OSHA Standards-Excavations, adopted March 5, 1990.

The contractor’s “responsible person” should evaluate the soil exposed in the

excavations as part of the contractor’s safety procedures. GROUND has provided the

information in this report solely as a service to the Interwest Consulting Group, and is not

assuming responsibility for construction site safety or the contractor’s activities.

The contractor should take care when making excavations not to compromise the

bearing or lateral support for any adjacent, existing improvements.

Unless analyzed individually, temporary, un-shored excavation slopes up to 15 feet in

height, in general, should be cut no steeper than 2 : 1 (horizontal : vertical) in the on-site

soils in the absence of seepage. Some surface sloughing may occur on the slope faces

at these angles. Should site constraints prohibit the use of the indicated slope angle,

temporary shoring should be used. GROUND is available to perform shoring design

upon request.

Groundwater and Surface Water Groundwater was encountered in the test holes as

shallowly as about 12 feet. Due to the proximity of the stream, the contractor should

anticipate encountering groundwater in project excavations advanced to depths greater

than about 10 feet. Water levels may be still shallower, however, at some times of the

year and will likely be similar to the stage of Big Thompson River. Wet soils likely will be

encountered above the actual water table.

Should seepage or flowing groundwater be encountered in project excavations, the

slopes should be flattened as necessary to maintain stability or a geotechnical engineer

should be retained to evaluate the conditions and provide additional or alternative

measures, as appropriate. The risk of slope instability will be significantly increased in

areas of seepage along excavation slopes. The contractor should be prepared to work

in wet conditions and in the presence of seepage. De-watering may be beneficial to

excavate and backfill project excavations.




The contractor should take pro-active measures to control surface waters during

construction and maintain good surface drainage conditions to direct waters away from

excavations and into appropriate drainage structures. A properly designed drainage

swale should be provided at the tops of the excavation slopes. In no case should water

be allowed to pond near project excavations.

Temporary slopes should also be protected against erosion. Erosion along the slopes

will result in sloughing and could lead to a slope failure.

PAVEMENT SECTIONS

A pavement section is a layered system designed to distribute concentrated traffic loads

to the subgrade. Performance of the pavement structure is directly related to the

physical properties of the subgrade soils and traffic loadings. The standard care of

practice in pavement design describes the flexible pavement section as a “20-year”

design pavement: however, most flexible pavements will not remain in satisfactory

condition without routine maintenance and rehabilitation procedures performed

throughout the life of the pavement. Pavement sections were developed in general

accordance with the design guidelines and procedures of the American Association of

State Highway and Transportation Officials (AASHTO), Colorado Department of

Transportation (CDOT), and Larimer County Urban Area Street Standards.

Subgrade Materials Based on the results of our field exploration and laboratory testing,

the potential pavement subgrade materials generally classify as A-2-4 to A-4 soil in

accordance with the AASHTO classification system with Group Index values of 0 to 1.

Resilient modulus testing was performed on a composite sample of shallow site soils

and yield a result of 11,146 psi at optimum moisture content. Based on our experience

at similar sites, however, significant variability in the resilient modulus of subgrade

materials should be anticipated along the project alignment. Therefore, to account for

variability across the alignment a resilient modulus values of 9,500 psi, corresponding to

an R-value of 40, was used to develop the pavement sections. It is important to note

that significant decreases in soil support have been observed as the moisture content

increases above the optimum. Pavements that are not properly drained may experience

a loss of the soil support and subsequent reduction in pavement life.




Anticipated Traffic The traffic data provided in the Project 318 Request for Proposal

and CDOT loading parameters were used to develop the anticipated project traffic.

These data indicated an average total of 9,400 vehicles per day over the design life of

the pavement, for the subject reach of Namaqua Road. Of these vehicles, 1.5 percent

are estimated by GROUND to be combination (semi-trailer) trucks, and 3.5 percent are

estimated by GROUND to be single-unit trucks. Based on these data, a design value for

flexible pavements of 1,147,464 ESAL18's was calculated (1,572,976 ESAL18's for rigid

pavements). These traffic loadings were used in design of the pavement section for

Namaqua Road.

If design traffic loadings differ significantly from the values above, GROUND should be

notified to re-evaluate the pavement section parameters below.

Pavement Sections The soil resilient modulus and the anticipated ESAL values were

used to determine the required structural number for the project pavements. The

required structural number was then used to develop minimum, pavement sections.

Pavement sections were based on the DARWin™ computer program that solves the

1993 AASHTO pavement design equation. A Reliability Level of 90 percent was utilized

develop the pavement sections. Structural coefficients of 0.44 and 0.11 were used for

hot bituminous asphalt and high quality aggregate base course, respectively. The

pavement section calculations are presented in Appendix B and the resultant minimum

pavement sections are tabulated below.

Minimum Pavement Sections

Subgrade R-Value Composite Section Rigid Section

40 6 inches Asphalt /

8 inches Aggregate Base8½ inches Concrete /

6 inches Aggregate Base

It should be noted that the calculated minimum pavement sections do not meet the

minimum county sections. We understand, however, that those minimums are for use in

the absence of geotechnical data. Therefore, the thinner, R-Value 40 sections may be

used.

Mill and Overlay Option It may be beneficial for the project to asphalt mill sections of

the alignment and overlay them with 2 or more inches of asphalt. Where asphalt milling




is selected along the alignment, the contractor should anticipate that local areas may

require more extensive repairs (i.e., the complete removal of a damaged section). Areas

needing full reconstruction should be identified in the field once the areas for mill and

overlay have been selected. Existing joints and cracking, along with other signs of

distress, commonly extend into the asphalt overlay after placement. However, with

routine, preventative maintenance, we estimate that the design life of Namaqua Road

can be extended up to 10 years where a 2 inch mill and overlay option is selected.

Thicker overlay sections may increase design life.

Pavement Materials Asphalt pavement should consist of a bituminous plant mix

composed of a mixture of aggregate and bituminous material. Asphalt mixture(s) should

meet the requirements of a job-mix formula established by a qualified engineer as well

as applicable Larimer County and CDOT design requirements.

Aggregate base material should meet the criteria of CDOT Class 6 aggregate base

course. Base course should be placed in and compacted in accordance with the criteria

in the Project Earthwork section of this report.

Where rigid (concrete) pavements are placed, the concrete should consist of a plant mix

composed of a mixture of aggregate, portland cement and appropriate admixtures

meeting the requirements of a job-mix formula established by a qualified engineer as

well as applicable design requirements of the Larimer County and CDOT. Concrete

should have a minimum modulus of rupture of third point loading of 650 psi. Normally,

concrete with a 28-day compressive strength of 4,500 psi should develop this modulus

of rupture value. The concrete should be air-entrained with approximately 6 percent air

and should have a minimum cement content of 6 sacks per cubic yard. Maximum

allowable slump should be 4 inches.

These concrete mix design criteria should be coordinated with other project

requirements including any criteria for sulfate resistance presented in the Water-Soluble

Sulfates section of this report. To reduce surficial spalling resulting from freeze-thaw

cycling, we suggest that pavement concrete meet the requirements of CDOT Class P

concrete. In addition, the use of de-icing salts on concrete pavements during the first

winter after construction will increase the likelihood of the development of scaling.

Placement of flatwork concrete during cold weather so that it is exposed to freeze-thaw

cycling before it is fully cured also increases its vulnerability to scaling. Concrete placing




during cold weather conditions should be blanketed or tented to allow full curing.

Depending on the weather conditions, this may result in 3 to 4 weeks of curing, or more.

Concrete pavements should contain sawed or formed joints. CDOT and various industry

groups provide guidelines for proper design and concrete construction and associated

jointing. The concrete pavement joints should be fully tied and doweled. Example

layouts that may be applicable for joints, as well as ties and dowels can be found in

CDOT’s M standards, found at the CDOT website:

http://www.dot.state.co.us/DesignSupport/. PCA and ACI publications also provide

useful guidance in these regards.

Subgrade Preparation Although subgrade preparation to a depth of 12 inches is typical

in the project area, pavement performance commonly can be improved by a greater

depth of moisture-density conditioning of the soils. The contractor should be prepared to

prepare the subgrade as outlined herein even where elevated subgrade moisture

contents are encountered beneath the existing pavements.

Remedial Earthwork Shortly before paving, the pavement subgrade should be

excavated and/or scarified to a depth of 12 inches or more, moisture-conditioned and

properly re-compacted. However, as noted in the Geotechnical Considerations for

Design section of this report, all existing, undocumented fill exposed soils in the

subgrade should be excavated and replaced with properly compacted fill. If the owner

opts to leave some of the existing fill soils in-place beneath paved areas, additional

settlements, accelerated pavement distress, and additional maintenance should be

anticipated. Similarly, where existing utility lines or other site constraints limit the depth

to which remedial earthwork can be accomplished, additional maintenance should be

anticipated.

Subgrade preparation should extend the full width of the pavement from back-of-curb to

back-of-curb. The subgrade for sidewalks and other project hardscaping also should be

prepared in the same manner.

Criteria and standards for fill placement and compaction are provided in the Project

Earthwork section of this report. The contractor should be prepared to either dry the

subgrade materials or to moisten them, as needed, prior to compaction.




Where adequate drainage cannot be achieved or maintained, excavation and

replacement should be undertaken to a greater depth, in addition to the edge drains

discussed below.

Proof Rolling Immediately prior to paving, the subgrade should be proof rolled with a

heavily loaded, pneumatic tired vehicle. Areas that show excessive deflection during

proof rolling should be excavated and replaced and/or stabilized. Areas allowed to pond

prior to paving will require significant re-working prior to proof-rolling. Establishment of a

firm paving platform (as indicated by proof rolling) is an additional requirement beyond

proper fill placement and compaction. It is possible for soils to be compacted within the

limits indicated in the Project Earthwork section of this report and fail proof rolling,

particularly in the upper range of specified moisture contents.

Additional Observations The collection and diversion of surface drainage away from

paved areas is extremely important to the satisfactory performance of the pavements.

The subsurface and surface drainage systems should be carefully designed to ensure

removal of the water from paved areas and subgrade soils. Allowing surface waters to

pond on pavements will cause premature pavement deterioration. Where topography,

site constraints, or other factors limit or preclude adequate surface drainage, pavements

should be provided with edge drains to reduce loss of subgrade support. The long-term

performance of the pavement also can be improved greatly by proper backfilling and

compaction behind curbs, gutters, and sidewalks so that ponding is not permitted and

water infiltration is reduced.

Landscape irrigation in planters adjacent to pavements and in “island” planters within

paved areas should be carefully controlled or differential heave and/or rutting of the

nearby pavements will result. Drip irrigation systems are suggested for such planters to

reduce over-spray and water infiltration beyond the planters. Enclosing the soil in the

planters with plastic liners and providing them with positive drainage also will reduce

differential moisture increases in the surrounding subgrade soils.

In our experience, infiltration from planters adjacent to pavements is a principal source of

moisture increase beneath those pavements. This wetting of the subgrade soils from

infiltrating irrigation commonly leads to loss of subgrade support for the pavement with

resultant accelerating distress, loss of pavement life and increased maintenance costs.

This is particularly the case in the later stages of project construction after landscaping




has been emplaced but heavy construction traffic has not ended. Heavy vehicle traffic

over wetted subgrade commonly results in rutting and pushing of flexible pavements,

and cracking of rigid pavements. In relatively flat areas where design drainage gradients

necessarily are small, subgrade settlement can obstruct proper drainage and yield

increased infiltration, exaggerated distress, etc. (These considerations apply to project

flatwork, as well.)

Also, GROUND’s experience indicates that longitudinal cracking is common in asphalt-

pavements generally parallel to the interface between the asphalt and concrete

structures such as curbs, gutters, or drain pans. Distress of this type is likely to occur

even where the subgrade has been prepared properly and the asphalt has been

compacted properly.

The anticipated traffic loading does not include excess loading conditions imposed by

heavy construction vehicles. Consequently, heavily loaded concrete, lumber, and

building material trucks can have a detrimental effect on the pavement.

Most pavements will not remain in satisfactory condition and achieve their “design lives”

without regular maintenance and rehabilitation procedures performed throughout the life

of the pavement. Maintenance and rehabilitation measures preserve, rather than

improve, the structural capacity of the pavement structure. Therefore, an effective

program of regular maintenance should be developed and implemented to seal cracks,

repair distressed areas, and perform thin overlays throughout the lives of the pavements.

The greatest benefit of pavement overlaying will be achieved by overlaying sound

pavements that exhibit little or no distress.

Crack sealing should be performed at least annually and a fog seal/chip seal program

should be performed on the pavements every 3 to 4 years. After approximately 8 to 10

years after construction, patching, additional crack sealing, and asphalt overlay may be

required. Prior to overlays, it is important that all cracks be sealed with a flexible,

rubberized crack sealant in order to reduce the potential for propagation of the crack

through the overlay. If actual traffic loadings exceed the values used for development of

the pavement sections, however, pavement maintenance measures will be needed on

an accelerated schedule.




CLOSURE

Geotechnical Review GROUND should review project plans and specifications to

evaluate whether they comply with the intent of the measures discussed in this report.

The author of this report or a GROUND principal should be contacted to provide that

review. The review should be requested in writing.

The geotechnical conclusions and parameters presented in this report are contingent

upon observation and testing of project earthworks by representatives of GROUND. If

another geotechnical consultant is selected to provide materials testing, then that

consultant must assume all responsibility for the geotechnical aspects of the project by

concurring in writing with the parameters in this report, or by providing alternative

parameters.

Materials Testing Interwest Consulting Group should consider retaining a geotechnical

engineer to perform materials testing during construction. The performance of such

testing or lack thereof, however, in no way alleviates the burden of the contractor or

subcontractor from constructing in a manner that conforms to applicable project

documents and industry standards. The contractor or pertinent subcontractor is

ultimately responsible for managing the quality of his work; furthermore, testing by the

geotechnical engineer does not preclude the contractor from obtaining or providing

whatever services that he deems necessary to complete the project in accordance with

applicable documents.

Limitations This report has been prepared for Interwest Consulting Group as it pertains

to design and construction of the proposed bridge and related improvements as

described herein. It may not contain sufficient information for other parties or other

purposes.

In addition, GROUND has assumed that project construction will commence by Spring,

2017. Any changes in project plans or schedule should be brought to the attention of a

geotechnical engineer, in order that the geotechnical conclusions in this report may be

re-evaluated and, as necessary, modified.

The geotechnical conclusions and criteria in this report relied upon subsurface

exploration at a limited number of exploration points, as shown in Figure 1, as well as

the means and methods described herein. Subsurface conditions were interpolated




between and extrapolated beyond these locations. It is not possible to guarantee the

subsurface conditions are as indicated in this report. Actual conditions encountered

during construction may differ from those encountered during site exploration.

If during construction, surface, soil, bedrock, or groundwater conditions appear to be at

variance with those described herein, a geotechnical engineer should be retained at

once, so that re-evaluation of the conclusions for this site may be made in a timely

manner. In addition, a contractor who relies upon this report for development of his

scope of work or cost estimates may find the geotechnical information in this report to be

inadequate for his purposes or find the geotechnical conditions described herein to be at

variance with his experience in the greater project area. The contractor is responsible

for obtaining the additional geotechnical information that is necessary to develop his

workscope and cost estimates with sufficient precision. This includes current depths to

groundwater, etc.

If any information referred to herein is not well understood, then Interwest Consulting

Group, the owner, or anyone using this report, should contact the author or a GROUND

principal immediately. We will be available to meet to discuss the risks and remedial

approaches presented in this report, as well as other potential approaches, upon

request.

This report was prepared in accordance with generally accepted soil and foundation

engineering practice in the Larimer County area at the date of preparation. Current

applicable codes may contain criteria regarding performance of structures and/or site

improvements which may differ from those provided herein. Our office should be

contacted regarding any apparent disparity. GROUND makes no warranties, either

expressed or implied, as to the professional data, opinions or conclusions contained

herein.

This document, together with the concepts and conclusions presented herein, as an

instrument of service, is intended only for the specific purpose and client for which it was

prepared. Reuse of or improper reliance on this document without written authorization

and adaption by GROUND Engineering Consultants, Inc., shall be without liability to

GROUND Engineering Consultants, Inc.




GROUND appreciates the opportunity to complete this portion of the project and

welcomes the opportunity to provide Interwest Consulting Group or the owner with a

proposal for construction observation and materials testing.

Sincerely,

GROUND Engineering Consultants, Inc.

Ben Fellbaum Reviewed by Brian H. Reck, P.G., C.E.G., P.E.

DIRECT SHEAR TEST REPORT

GROUND ENGINEERING CONSULTANTS, INC.ENGLEWOOD, CO.

Client: Interwest Consulting Group

Project: Replacement Namaqua Avenue Bridge Replacement over the Big

Thompson River

Location: TH-2

Depth: 17 feet

Proj. No.: 16-3048 Date Sampled: 8/15/16

Sample Type: Liner

Description: Silty SAND

LL= NV PI= NP

Specific Gravity= 2.69

Remarks:

Figure 8

Sample No.

Water Content, %

Dry Density, pcf

Saturation, %

Void Ratio

Diameter, in.

Height, in.

Water Content, %

Dry Density, pcf

Saturation, %

Void Ratio

Diameter, in.

Height, in.

Normal Stress, psi

Fail. Stress, psi

Strain, %

Ult. Stress, psi

Strain, %

Strain rate, in./min.

Initia

lA

t T

est

Sh

ea

r S

tre

ss,

psi

0

10

20

30

40

50

60

Strain, %

0 5 10 15 20

1

2

3

Fail.

Str

ess,

psi

0

20

40

60

Normal Stress, psi

0 20 40 60 80 100 120

C, psi

, deg

Tan()

Results

4.31

33.2

0.65

1

11.6

105.1

52.0

0.5973

1.94

1.00

7.7

106.2

35.6

0.5819

1.94

0.9910.0010.72

3.6

0.01

2

11.6

105.7

52.8

0.5885

1.94

1.00

7.7

108.1

37.2

0.5541

1.94

0.9820.0017.59

6.2

0.01

3

11.6

26.4

5.8

5.3542

1.94

4.00

7.3

26.6

3.7

5.3086

1.94

3.9740.0030.41

8.8

0.01

0.5

0.6

0.7

0.8

0.9

1

1.1

1 1.5 2 2.5 3 3.5

Center to Center Spacing (Pier / Pile Diameter)

Ax

ial

Ca

pa

cit

yR

ed

uc

tio

n F

ac

tor

-0.100

0.000

0.100

0.200

0.300

0.400

0.500

0.600

0.700

0.800

0.900

1.000

1.100

0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0 4.5 5.0 5.5 6.0 6.5 7.0 7.5 8.0 8.5 9.0

1st Pier / Pile

2nd Pier / Pile

3rd & Subsequent Piers / Piles

Center to Center Spacing (Pier / Pile Diameter)

La

tera

l C

ap

ac

ity

Re

du

cti

on

Fa

cto

r (p

mu

ltip

lier)

TABLE 1

SUMMARY OF LABORATORY TEST RESULTS

Sample Location Natural Natural Percent Atterberg Limits Percent Unconfined USCS AASHTO

Test Moisture Dry Passing Liquid Plasticity Swell Compressive Classifi- Classifi- Soil or

Hole Depth Content Density Gravel Sand No. 200 Limit Index (Surcharge Strength cation cation Bedrock Type

No. (feet) (%) (pcf) (%) (%) Sieve Pressure) (psf) (GI)

1 24 14.9 SD - - 10 21 5 - - SP-SC A-2-4 (0) SAND with Silty, Clay

2 7 4.7 SD - - 14 NV NP - - SM A-2-4 (0) FILL: Silty Sand

2 17 11.6 SD - - 13 NV NP - - SM A-2-4 (0) Silty SAND

2 30 6.0 120.3 - - 91 29 8 - 22,360 CL A-2-4 (0) Clay SHALE Bedrock


2 - 6.1 SD - - 87 27 7 - - CL A-2-4 (0) Clay SHALE Bedrock

3 5 19.3 96.5 - - 75 25 3 - - ML A-4 (1) FILL: Silt with Sand

3 15 14.8 - 1 79 20 - - - - SM - Silty SAND

3 30 9.7 124.2 - - 90 31 11 - 11,620 CL A-6 (1) Clay SHALE Bedrock

4 3 8.4 101.2 - - 19 NV NP - - SM A-2-4 (0) FILL: Silty Sand

4 13 11.0 106.8 4 88 8 NV NP - - SP-SM A-2-4 (0) Silty SAND

4 29 7.3 113.3 - - 92 30 8 - - CL A-2-4 (0) Clay SHALE Bedrock


P-1 1 6.0 122.9 - - 18 NV NP - - SM A-2-4 (0) FILL: Silty Sand

P-2 3 16.3 102.9 - - 39 23 4 -0.1 (150 psf) - SC-SM A-4 (0) FILL: Silty, Clayey Sand

P-3 2 11.5 105.9 - - 40 21 3 - - SM A-4 (0) FILL: Silty Sand

P-4 2 14.9 111.8 - - 47 25 5 - - SC-SM A-4 (0) FILL: Silty, Clayey Sand

P-5 3 18.9 98.0 - - 47 24 3 -0.2 (400 psf) - SM A-4 (0) FILL: Silty Sand

P-6 1 5.7 99.2 - - 15 20 2 - - SM A-2-4 (0) FILL: Silty Sand

Job No. 16-3048

Gradation

TABLE 2

SUMMARY OF SOIL CORROSION TEST RESULTS

Sample Location Water Redox Sulfides USCS

Test Soluble pH Potential Content Resistivity Classifi- Soil or

Hole Depth Sulfates cation Bedrock Type

No. (feet) (%) (mV) (ohm-cm)

3 5 0.46 8.6 -95 Positive 953 ML SILT with Sand

P-2 3 0.02 - - - - SM FILL: Silty Sand

P-6 1 0.06 - - - - SM FILL: Silty Sand

Job No. 16-3048

APPENDIX A

Core Hole Logs

APPENDIX B

Pavement Section Calculations

Page 1

1993 AASHTO Pavement Design

DARWin Pavement Design and Analysis System

A Proprietary AASHTOWareComputer Software Product

Network Administrator

Flexible Structural Design Module

16-3048Namaqua Bridge ReplacementAsphalt - Composite Section

R-40

Flexible Structural Design

18-kip ESALs Over Initial Performance Period 1,147,464 Initial Serviceability 4.5 Terminal Serviceability 2.5 Reliability Level 90 %Overall Standard Deviation 0.44 Roadbed Soil Resilient Modulus 9,500 psiStage Construction 1

Calculated Design Structural Number 3.15 in

Specified Layer Design

Layer Material Description

StructCoef.(Ai)

DrainCoef.(Mi)

Thickness(Di)(in)

Width(ft)

CalculatedSN (in)

1 Asphalt 0.44 1 6 - 2.642 Road Base (CDOT Class 6) 0.11 1 8 - 0.88

Total - - - 14.00 - 3.52

Page 1

1993 AASHTO Pavement Design

DARWin Pavement Design and Analysis System

A Proprietary AASHTOWareComputer Software Product

Network Administrator

Flexible Structural Design Module

16-3048Namaqua Bridge ReplacementAsphalt - Full Depth Section

R-40

Flexible Structural Design

18-kip ESALs Over Initial Performance Period 1,147,464 Initial Serviceability 4.5 Terminal Serviceability 2.5 Reliability Level 90 %Overall Standard Deviation 0.44 Roadbed Soil Resilient Modulus 9,500 psiStage Construction 1

Calculated Design Structural Number 3.15 in

Specified Layer Design

Layer Material Description

StructCoef.(Ai)

DrainCoef.(Mi)

Thickness(Di)(in)

Width(ft)

CalculatedSN (in)

1 Asphalt 0.44 1 7.5 - 3.30Total - - - 7.50 - 3.30

Rigid Pavement Design - Based on AASHTO Supplemental Guide

Reference: LTPP DATA ANALYSIS - Phase I: Validation of Guidelines for k-Value Selection and Concrete Pavement Performance Prediction

Results

Project # 16-3048Description: Namaqua Bridge Replacement

Location: Loveland, CO

Slab Thickness Design

Pavement Type JPCP18-kip ESALs Over Initial Performance Period (million) 1.60 millionInitial Serviceability 4.5Terminal Serviceability 2.528-day Mean PCC Modulus of Rupture 650 psiElastic Modulus of Slab 3,400,000 psiElastic Modulus of Base 30,000 psiBase Thickness 6.0 in.Mean Effective k-Value 25 psi/inReliability Level 90 %Overall Standard Deviation 0.34

Calculated Design Thickness 7.85 in

Temperature Differential

Mean Annual Wind Speed 8.8 mph

Mean Annual Air Temperature 48 oFMean Annual Precipitation 15.7 in

Maximum Positive Temperature Differential 6.02 oF

Modulus of Subgrade Reaction

Period Description Subgrade k-Value, psi

3 Winter 253 Spring 253 Summer 253 Fall 25

Seasonally Adjusted Modulus of Subgrade Reaction 25 psi/in

Modulus of Subgrade Reaction Adjusted for Rigid Layer

and Fill Section 25 psi/in

Traffic

Performance Period yearsTwo-Way ADTNumber of Lanes in Design DirectionPercent of All Trucks in Design LanePercent Trucks in Design Direction

Vehicle Class Percent of Annual Initial Annual AccumulatedADT Growth Truck Factor Growth in 18-kip ESALs

Truck Factor (millions)

Total Calculated Cumulative ESALs million

Faulting

Doweled

Dowel Diameter 1 inDrainage Coefficient 1.00

Average Fault for Design Years with Design Inputs 0.05 inCriteria Check PASS

Nondoweled

Drainage Coefficient 1

Average Fault for Design Years with Design Inputs inCriteria Check

Traffic Loading Calculations GROUND Job No. 16-3048

18 kip Loadings, S.I. = 2.5 Equivalent 18 kip Axle Loadings

High Average Method

Alignment: Namaqua Bridge

ESAL18's

Long Term Traffic Flexible Rigid

Pavement Pavement

Total Vehicles per day: 9,400

% Combination Trucks: 1.5 % 141 vehicles 671,309 1,044,945

% Single Unit Trucks: 3.5 % 329 vehicles 358,814 410,691

% Cars & Pickups: 95 % 8930 vehicles 117,340 117,340

No. Of Traffic Lanes = 2 Total: 1,147,464 Total: 1,572,976

(total, both ways) Lane Factor: 0.60 EDLA: 157 EDLA: 215

Design Life (Years): 20 Source of Traffic Data: Larimer County RFP for Project 318

Geotechnical Evaluation Replacement Namaqua Bridge Over the … · 2018-11-28 · Typical Heel Drain Detail ... abutment and/or wing walls. Additionally 300 to 400 feet of Namaqua

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