ASCE - Mexico Viaduct Field Test

8/20/2019 ASCE - Mexico Viaduct Field Test

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Elevated Viaducts Tested

The superstructure of all three elevated viaducts under study was in-tegrated by several precast and cast-in-place concrete elements.A typical section of the elevated viaducts consists of a series of Gerber girders in which a simply supported girder (TC) rests on the projectingcantilevers of two adjacent girders (TA), which are in turn connected toa pair of cantilevered columns forming a rigid frame in the longitudinaldirection. Fig. 2 shows the main geometric properties of the superstruc-

ture in sites PI, PV1, PV2, and VB. The clear span at the four sitesselected for the tests ranged from 35 to 43 m. The depth of the girderswent from 2.0 to 2.5 m, and thewidth of the superstructure ranged from 9.15 to 12.65 m. The clear height of the columns in the test sites rangedfrom 10 to 17 m (Fig. 3). The roadway at all test locations was threelanes wide, with the exception of site PI, which had two lanes. Exclud-ing the box girders at test site PV1, these structural elements had sym-metric flanges a test sites PI, PV2, and VB. The connection between

columns and girders was cast in place after running the prestressingstrands of the columns through ad hoc openings left in the girdersand the placement of additional shear and confining reinforcement in the joint. The monolithic connection was later posttensioned (Fig. 4).

The foundation for each of the supporting columns consisted of a single footing connected to four cast-in-place reinforced concretepiles. Structural configurations proposed by the designer in sitesPI, PV1, PV2, and VB had slight differences. The structure in sitePI had reinforced concrete piers supporting large prestressed con-

crete girders [Fig. 3(a)]. The configuration at sites PV1 and PV2,both of which were located along the same elevated viaduct, con-sisted of prestressed concrete columns, girders, and reinforced con-crete decks [Fig. 2(c)]. For site VB, the design firm changed from theoriginal design of sites PV1 and PV2 to a system of prestressed con-crete box girders and reinforced concrete decks resting on pre-stressed pier-footing assemblies [Fig. 2(d)]. This decision wasbased on the preliminary results of the load tests at sites PV1 and

Fig. 1. (Color) View of the elevated viaduct during construction and field tests in site VB: (a) general view of the site; (b) lateral load tests;

(c) vehicular load tests (images by Instituto de Ingeniería, UNAM)

Fig. 2. (Color) Structural configuration of elevated viaducts tested: (a) typical elevation of the superstructure; (b) site PI; (c) site PV1 and PV2;

(d) site VB

© ASCE D4014001-2 J. Struct. Eng.

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PV2. The posttensioning of all joints in the structure allowed for theconnections to be considered monolithic in terms of stiffness andstrength. Figs. 4 and 5 show relevant details of the reinforcement in the column-girder, column-footing, and footing-pile connectionsin the test sites.

Test Site PI

The selected area of site PI was composed of two longitudinalframes connected by means of a TC girder. Fig. 2(b) shows thegeneral configuration of the superstructure at site PI. All columnsin the test site had the same 1.10× 1.50 m oval cross sectionthrough their height, and were founded on 2.00-m-deep, 3.75×

3.75 m square footings. Each footing was supported on four 12.50-m-long circular piles with a diameter of 0.60 m. The clear height of the columns was in the range of 10.12–11.20 m, and themain span of the deck was 35.00 m. The box girders were 2.00 m deep, 9.15 m wide, and had symmetric flanges. The roadway ac-commodated two traffic lanes.

The concrete compressive strength used for the design of theprestressed box girders was 59 MPa. The compressive strength of

concrete for columns and footings was 44 MPa. The nominal val-ues for the yield strength of prestressing steel and reinforcing steelwere 1,862 and 412 MPa, respectively (Riobóo and Meli 2003).

Test site PI is located in the so-called firm zone of the seismicmap for the metropolitan area of Mexico City. Structures in thiszone are designed with a seismic coefficient of 0.16, which maybe increased 50% to account for the importance of the structure.

Fig. 3. Columns dimensions in test sites: (a) site PI; (b) site PV1; (c) site PV2; (d) site VB

Fig. 4. Typical column-girder connection in site VB (other sites are

similar)

Fig. 5. Typical footing-column connection in (a) sites PV1 and PV2; (b) site VB


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The dominant frequency of the soil at site PI, determined throughambient vibration tests, was in excess of 1.42 Hz.

Test Sites PV1 and PV2

Sites PV1 and PV2 had precast columns with a 1.80× 2.40 m and 1.50× 2.20 m oval cross section, respectively. Cross sectionsof columns in both sites were the same along their height. Thelower 2.00-m-long portion of the columns was inserted in a 3.95× 5.75 m, chandelier-like, reinforced concrete footing with

a depth of 2.65 m [Fig. 5(a)]. The gap left between the insertedcolumns and the footing was filled with a high-fluidity, high-strength mortar, and then, on the top surface of the footing, a post-tensioned concrete collar was cast around the perimeter of thecolumn in order to increase the overall stiffnesses of the assembly[Fig. 5(a)]. This type of foundation is different from the socket connection referred to by Haraldsson et al. (2013) and Osanai et al.(1996), which requires precise dimensioning and detailing of thecolumn-footing joint. Such precision was not attainable at the fab-rication sites, which led to the choosing of a posttensioned system despite being somewhat difficult to execute and thus requiringcloser supervision.

Similarly to test site PI, each footing was supported by four 16.50-m-long circular piles with 0.80 m in diameter. The clear

height of the columns in sites PV1 and PV2 was 14.20 and 13.60 m,respectively. The superstructure was composed of a TC girder,which rested as a simply supported member at the ends of twocantilevered TA girders. The main span of the deck at both siteswas 43.00 m. In both cases, the box girders were 2.40 m deep and12.65 m wide, accommodating three traffic lanes. The flanges at opposite sides of the girders were asymmetric at site PV1 and sym-metric at site PV2 [Figs. 3(b and c)].

The concrete compressive strength used for the design of theprestressed box girders and the columns was 59 MPa. The com-pressive strength of concrete for the footings was 44 MPa. Themechanical properties of the prestressing steel and reinforcing steelin sites PV1 and PV2 were the same as those for test site PI. SitesPV1 and PV2 were located in the same seismic zone as site PI.

The dominant frequency of the soil at both sites PV1 and PV2was determined through ambient vibration tests and was foundto be in excess of 1.4 Hz.

Test Site VB

In site VB, the test region included four columns identified with themarks A, B, C, and D and whose clear heights were 11.00, 10.75,9.75, and 9.25 m, respectively. The general configuration of thesuperstructure and foundation at site VB is shown in Figs. 2–5.In this case, the columns and their footings were prefabricatedand cast integrally at the prestressing plant. The rectangular cross-section of the hollow columns changed over their height. Columndimensions were 1.83× 2.50 m at their base, 1.40× 2.50 m at their cross-section, with a 3.40× 2.50 m rectangular capital at their top. The thickness of the walls on the hollow columns was approx-imately 0.40 m.

Each footing was 4.60× 3.60 m in plan and 1.70 m deep. Thelonger side of the footing was perpendicular to the direction of traf-fic in the elevated viaduct. The footings rested on a 0.25-m-thick reinforced concrete slab and a 1.70-m-thick flowable fill. The sametype of fill was provided around the footing to reach a depth of 2.20 m. The overall dimension of the footing and the surroundingflowable fill in plan view was around 7.60× 6.60 m. Four 0.90-m-diameter perforations were allocated in each footing to provideproper anchorage and development of the piles reinforcing steel

and achieve a monolithic connection between the footing and thereinforced concrete piles previously cast in situ. The length of thepiles ranged from 22.60 to 24.10 m, and their diameter was 0.80 m.The joints between the reinforced concrete piles of the foundationand the pier-footing assembly, and between said assembly and thebox girders, were posttensioned in situ [Fig. 5(b)].

The superstructure of the elevated viaduct at test site VB wascomposed of a simply supported TC girder, which rested at the endsof two cantilevered TA girders. Each of the TA girders rested ona pair of columns forming monolithic frames in the longitudinal

direction of the viaduct. The clear span was 34.00 m and the totallength of the TA girder was 25.00 m. The box girders had a depthof 2.40 m and a width of 12.35 m, adequate for three traffic lanes.

Nominal compressive strength of concrete for all structuralelements was 59 MPa. The nominal yield strength of prestressingsteel was 1,862 MPa, whereas the nominal yield strength of reinforcing steel was 412 MPa.

The soil at site VB was characterized by the existence of sandylayers up to a depth of 10 m, and the presence of sand layerswithin depths of 10 to 32 m. The shear wave velocity of theselayers has been measured to range from 100 to 300 m =s. The firm deposits underneath the sand layers have shear wave velocities of around 400 m =s. The dominant frequency of the soil at site VBwas determined through ambient vibration tests and was found to

be around 1.08 Hz. The use of the approximate formula providedin the seismic design code of Mexico City for soft soils( f ¼ Vs=4H , where f is the dominant frequency, Vs is the shear wave velocity of the medium, and H is the thickness of themedium assumed as semi-infinite) results in a dominant frequencyof 1.6 Hz. The use of this expression, however, was deemed in-appropriate since site VB lies precisely on the limit between thetransition and firm zones of the seismic zoning map of MexicoCity. The geotechnical surveys conducted by Mooser et al.(1996) and Juárez et al. (2010) indicate that site VB is locatedin an alluvial fan, making it a geotechnical singularity in whichthe simplifications of the approximate formula cannot be assumed.Structures in this zone must be designed with a seismic coefficient of 0.32. A 50% increment of this value was enforced for the de-

sign of the viaduct as specified for important structures in the de-sign code and standards for Mexico City.

Testing Programs

The field test programs included three types of tests: lateral loadtests in which a pulling lateral load of increasing magnitude wasapplied at the top of the columns during construction and whenthe project was concluded [Fig. 1(b)]; vehicular load tests withthe use of loaded/unloaded and parked/moving trucks [Fig 1(c)];and ambient vibration tests. A different set of tests was used for each site depending on the availability of space, time, and resour-ces. Table 1 presents the testing program matrix for all test sites.

All field tests were carried out during nighttime and before theopening of the viaducts to public service. The instrumentation at each one of the sites had to be envisioned under numerous limita-tions imposed by the construction companies and their subcontrac-tors so that it would not interfere with their progress. As a result, theinstrumentation scheme was modified and improved from one test site to the next (Fig. 6).

Lateral Load Tests

The lateral load tests were carried out in sites PV1, PV2, and VB.The tests consisted of applying an increasing lateral load at the topof a single column by means of a 4,900-kN crane. The load was


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applied perpendicularly to the direction of traffic and as parallel toground level as possible. A pair of ad hoc steel plates was designed

and embedded in the concrete at the top of the columns to allow for the application of the load. A steel cable was used to pull the col-umn by linking the embedded steel plates to the hook of the crane.The horizontality of the pulling cable was monitored by means of surveying equipment. The main purpose of the lateral load tests wasto determine the lateral displacement of the column under cumu-lative loads, with a focus on the overall column-footing behavior.The magnitude of the lateral load was similar to the value corre-sponding to the calculated cracking moment, and the column’sbehavior was kept in the cracked elastic range. In site PV1, thelateral load associated with the calculated cracking moment of the column was applied. For site PV2, the maximum applied lateralload was that corresponding to the cracking moment. In both sites,

the cracking moment was computed on the basis of the calculatedmodulus of rupture (0.7p f 0c < f t


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Displacement transducers were fixed to an independent steelstructure erected in the vicinity of the columns, allowing for themeasurement of their lateral displacement. In sites PV1 and PV2,located in areas of firm soils, the foundation of the steel frame was1 m deep. An existing 3-m-high slope supported the frame laterally.In site VB, a rigid steel structure was used. The fundamental fre-quency of the steel frame was measured to be more than 6.5 timesthat of the soil at the site. The local rotation at the base of the steelframe was monitored during the tests and measured as 7 × 10−6 radwhen the largest lateral loads were applied. Such a rotation at the

base resulted in differences between measured and actual values for the lateral displacement that were under 4.5%. The steel structureserved as an isolated frame of reference and was designed withenough rigidity so that any lateral deformation was negligible.Potentiometers and inclinometers were mounted directly on thesurface of the columns by means of drilled anchors and aluminum plates. The use of surveying equipment made possible to corrobo-rate the measured lateral displacements not only for the columnsubjected to lateral load but also in the two adjacent columns.The instrumentation schemes depicted in Fig. 6 and the surveyingdata acquired allowed for an accurate description of the deformedshape of the columns under various load conditions as well as for the computation of the rotation at their bases.

Vehicular Load Tests

Static and dynamic vehicular load tests were carried out in sites PA,PV1, and VB. The vertical loads were imposed by means of T3-S3and T3-S2 trucks. Four loaded trucks with an average weight of 573 kN and a single truck of 196 kN were used for the tests at sitesPA and PV1. Five trucks with an average weight of 598 kN anda 160-kN truck were used at site VB. The purpose of these testswas to determine the lateral and vertical displacements of the super-structure under static, dynamic, and impact loads in order to com-pare them with the design values.

For the static vehicular tests, different arrays of parked truckswere envisioned. For the dynamic vehicular tests, the sametrucks were driven at various speeds along the central lane of the

elevated viaducts. In order to evaluate the influence of wearing-surface irregularities and to produce additional impact loads, someof the dynamic vehicular tests were run after setting a 50-mm-high,500-mm-long, and 3-m-wide steel ramp, which served as a tempo-rary speed bump on the asphalt wearing surface (Cantieni 1983;Bank and Pinjarkar 1989; Paultre et al. 1992). The run of the trucksover the speed bump without breaking replicated the occurrenceof short-duration concentrated loads, commonly referred to as im-pact loads.

Accelerations and vertical displacements of the superstructurewere recorded during the vehicular tests using servo-accelerometersand displacement transducers located at strategic points of the struc-ture. Instrumentation at sites PI and PV1 included 9 accelerometersand 23 displacement transducers. The instrumentation deployment

at site VB comprised 26 accelerometers and 33 displacement trans-ducers. Accelerometers were mounted directly on the superstructureby means of aluminum plates. The displacement transducers ranfrom different points of the superstructure to the ground by meansof steel pipes. This instrumentation scheme permitted the measure-ment of the vertical deflection of girders when subjected to variousload conditions.

Ambient Vibration Tests

Ambient vibration tests were carried out at all four test sites. Thesetests were conducted before and after any other tests. The purpose

of the ambient vibration tests was to determine the natural frequen-cies of the structure and the soil, and to assess if the lateral load testsor the vehicular load tests had any impact in the original dynamicproperties of the structure. Various arrays of servo-accelerometerswere deployed during the ambient vibration tests.

Test Results

Lateral Load Tests in Sites PV1 and PV2

One column at each of test sites PV1 and PV2 was selected toevaluate the behavior of the cantilevered element under lateral loadsand the effect of the posttensioned collar slab on the behavior of thecolumn-footing ensemble. Due to the height of the columns andthe space restraints around them, the lateral displacement wasmeasured using displacement transducers (LVDTs) over the lowest 6.8 m of their height. The lateral displacements at higher columnlocations were determined with topographic equipment. When at-tempting to match the lateral displacements recorded by the LVDTsto the data obtained with the electronic surveying station, a largescatter was noticed in the survey data points. Thus, a methodologywas proposed to determine the deformed shape of the column onthe basis of all LVDT measurements in the lower portion of the

column and a single topographic data point corresponding to thetop of the column. Displacement data points were adjusted by sub-tracting the displacement measured at the footing, and their valuescompared with those of a simple model based on the deformedshape of a prismatic, perfectly elastic, homogeneous, and weight-less cantilevered element subjected to a lateral load at its highest point. Even though the weight of the superstructure and the columnmay be quite large and generate permanent deformations, the pro-cedure focuses on determining the deformation solely due to theapplication of the externally applied lateral load while ignoringany previous deformations due to permanent loads. If the columnheight is H and the load applied is P, the lateral displacement of a cross-section located at a distance y from the base of the columnis given by Eq. (1)

x ¼ P6EI

ðy3 − 3H 2Þ ð1Þ

However, the conditions in the field are somewhat different since there is self-weight, structural redundancy at the top of thecolumn and the fixation at its base might not perfect as assumedfor the development of Eq. (1). In order to account for these dis-crepancies, it is assumed that only a fraction (K ) of the lateral loadis required in the actual column to produce the same displacement as in the model

x ¼ KP6EI

ðy3 − 3H 2Þ ð2Þ

Defining C as

C ¼ K 6EI

ð3Þ

Substituting Eq. (3) in Eq. (2), and solving for C results in

C ¼ xðy3 − 3H 2ÞP ð4Þ

By using Eq. (4), the values for C and their statistical modeswere calculated for each load step. In this process, it is assumedthat the product EI remained constant during the lateral load tests.The lateral displacement of the columns obtained after adjusting the


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surveyed data by using the described methodology were consistent with the displacements measured by the transducers. The valuescalculated for C using Eq. (4) suggest that any relative movement between the column and its footing did not have a noticeable effect on the elastic response of the ensemble. It must be noted that themain purpose of this methodology is to relate measured lateral dis-placements with those provided by a simplified structural model.

The lateral displacements of adjacent columns were measured at sites PV1 and PV2. Throughout the tests, the displacementsobserved in neighboring columns remained under 50% of those

measured for the column under direct lateral loading. The former is indicative that a portion of the lateral load was transferred toadjacent members through the participation of the box girders inthe load-resisting mechanism.

During the before-collar tests, the first flexural crack incolumn B was observed when the lateral load was increased from 382 to 490 kN. Considering the variation of the maximum tensilestress for the transition between uncracked and cracked sectionsfor Class T prestressed members (Table R18.3.3 of ACI 318-11;ACI 2011), the lateral load associated with the cracking moment was calculated to be between 500 and 675 kN. The first flexuralcrack was measured to be 0.07 mm wide and 200 mm long whenthe lateral load had been increased to 549 kN. This crack was not visible after the removal of the lateral load. During the second lat-

eral load cycle, the flexural crack was visible again when the lateralload reached 382 kN. Additional flexural cracks located in the low-est 1.5-m-long region of the column appeared as the lateral loadreached 637 and 695 kN. After the lateral load was removed, someflexural cracks remained visible. The maximum width of flexuralcracks present after the load tests was 0.10 mm. The width of thecracks that remained visible was under the limit set in the code for the service load condition. Even though the presence of thesecracks was undesirable, they are unlikely to affect the structuralresponse of the members under service and ultimate loads. It must be noted that the assumption of the EI product to be constant for thederivation of Eq. (4) would not be true after flexural cracking hasdeveloped. However, for the target loads of the tests, the errorsinduced by this discrepancy were under 4%.

After the before-collar tests were concluded, an additionaldisplacement transducer was placed in the region where flexuralcracking had concentrated (1.24 m from the base of the column)in order to estimate average deformations in that region. The rela-tionship between the average strains at the base of the column mea-sured during the application of the three cycles of lateral load, thelateral load, and the equivalent moment are shown in Fig. 7. Thedeformations are compared with strains computed on the basisof the classic Euler –Bernoulli beam theory. The measured strainsagreed well with those calculated. Fig. 7 shows similar trends for all load cycles, including hysteretic behavior associated with re-duction in the axial compressive stresses for lateral loads larger than 180 kN. The axial decompression was due to the presenceof tensile stresses induced by the flexural moment at the base of the column.

During the after-collar tests, the width and length of the flexuralcracks that had appeared during the before-collar tests increased.Nevertheless, the width of flexural cracks was less than 0.10 mm for lateral loads up to 695 kN. For all before-collar and after-collar tests, no cracks were visible in any of the adjacent columns. Despitethe appearance of cracks during the load tests, no repair or reha-bilitation was deemed necessary.

When comparing the lateral displacement of the columns for thebefore-collar and after-collar condition at site PV2, it was observedthat they were practically identical for lateral forces under 206 kN.For larger lateral forces, however, the lateral displacements of the

column in the after-collar condition were consistently smaller that

those corresponding to the before-collar condition (Fig. 8). It wasevident that the cast of the posttensioned collar slab resulted in anincrease of the lateral stiffness of the structure, thus reducing itslateral displacement.

For both the before-collar and after-collar tests, the lateral dis-placements measured at the base of column B were consistentlylarger than those measured at the footing. This behavior couldbe associated with the presence of the grout-filled joint betweenthe column and the footing, which induced a strain concentration.However, after the posttensioned collar was in place, the displace-ments measured at the grout-filled joint were reduced 38% withrespect to those measured during the before-collar tests. Moreover,the measured angular rotations at the bend of column B were com-pared with the calculated rotations assuming a cantilevered column.

The results consistently showed that the translational and rotationalrestraints imposed by the superstructure should not be ignored.

Lateral Load Tests in Site VB

In site VB, the column marked with letter B in Figs. 1(a) and 3(b)was subjected to monotonically increasing lateral loads. The testswere conducted at two stages of construction. Two target loads (Ti)were set for the tests during the first stage of construction (canti-levered column) and four target loads (Tc) were established for the tests of the second stage (construction completed). The lateraldisplacement at various points along the height of column B was

Fig. 7. (Color) Applied load versus strain measured for the after-collar

condition of column B in site PV2

Fig. 8. (Color) Measured and calculated lateral displacements for

column B in site PV2


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measured by potentiometers. Fig. 9 shows the data points obtainedfor the effective horizontal load tests carried out for both stages of construction. Fig. 9(a) shows the approximate lateral deflectionshape of column B for target loads of 230 and 431 kN.

It is evident that all lateral displacements reduced from the first to the second stage tests. Indeed, the presence of the superstructureresulted in a reduction of all lateral displacements of column B bya factor of around 4. Fig. 9(b) presents the results of Fig. 9(a) after a normalization of the lateral displacements was done. The coinci-dence of all data points indicates that the deformed shape of column B did not change from the first to the second stage tests.The presence of the superstructure did not modify the distributionof lateral displacements over the height of the columns even if issignificantly reduced their magnitude. However, the local deforma-tion at the base of column B, inferred by both lateral displacementsand rotations in that zone, did change from the tests of the first stageto those of the second stage.

The instrumentation scheme deployed at site VB allowed for the estimation of the static translational and rocking stiffnesses as-sociated with soil-structure interaction (SSI) in the transverse direc-

tion of the viaduct. The first-stage test data, corresponding to thecantilevered column, was used for the estimation. Fig. 10 presentsthe lateral load versus base displacement and the moment versusbase rotation plots recorded for the two target effective hori-zontal loads applied during the first stage tests (Ti1 ¼ 230 kNand Ti2 ¼ 555 kN). The translational stiffnesses were obtained asthe slope of linear fits to the data points in Fig. 10(a). A single linewas enough to describe the quasi-linear behavior of the test points

corresponding to the first target lateral load (Ti1). For the secondtarget lateral load (Ti2), a trilinear fit described the test points

best. The cutoff points for the trilinear fit were identified at 147and 343 kN.For estimating the rocking stiffnesses of the SSI system at site

VB, the plots in Fig. 10(b) were used. For the lateral load test tothe first target load (Ti1), a single line was sufficient to represent the quasi-linear distribution of the moment versus base rotation data points. For the test to the second target load (Ti2), a singularity wasobserved for a moment of around 3,430 kN · m, corresponding to aneffective horizontal load of around 343 kN. Note that the same linefits the data points for both target load tests up to a moment of 3,000 kN · m, associated with a lateral load of 300 kN. Also notethat the onset of the second line fit for the moment versus base ro-tation data points (3,430 kN · m) coincides with the onset of thethird segment of the line fit for the effective horizontal load versus

lateral deflection data points (343 kN). The values for the static trans-lational and rocking stiffnesses obtained from the linear fit of the test data of column B in test site VB are summarized in Table 2.

Static Vehicular Load Tests

The static vehicular load tests consisted of parking a number of loaded and unloaded trucks on the superstructure. Four arrays for the location of trucks were used in sites PI and PV1 [Fig. 11(a)].Five arrays were used in site VB [Fig. 11(b)]. The arrays at eachsite were designed on the basis of producing both the maximum vertical displacements at the midspan cross-section of the box gird-ers (TC), and the largest lateral displacements of the instrumentedcolumns (B).

The experimental data corresponding to the vertical displace-ments at various points of the superstructure were used to plot itsapproximate deformed shape at selected cross-sections of the test sites under various load conditions. Fig. 8 shows the variation of the vertical displacements measured at sites PV1 and VB in threecross-sections of the superstructure. The three lines shown for eachplot in Fig. 12 correspond to the approximate deformed shape at thecross-sections identified as A, B, and Midspan in Fig. 11. Note that

Fig. 9. (Color) Measured lateral displacement for column B in site VB:

(a) lateral displacement; (b) normalized lateral displacement

Fig. 10. (Color) Estimation of translation and rocking stiffnesses in site VB: (a) lateral load versus base displacement; (b) moment versus base

rotation

Table 2. Estimated Static Translational and Rocking Stiffnesses in Site VB

Stiffness

Effective horizontal load

0–147 kN 147–343 kN 343–555 kN

Translational, in MN=m 4,528 882 412Rocking, in MN · m =rad 46,746 46,746 33,812


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the vertical displacements for Arrays IV and V in site VB arepresented in the same plot.

For the lateral load test to the first target load (Ti1), a singleline was sufficient to represent the quasi-linear distribution of themoment versus base rotation data points. For the test to the secondtarget load (Ti2), a singularity was observed for a moment of around 3,430 kN · m, corresponding to a lateral load of around343 kN. Note that the same line fits the data points for both target

load tests up to a moment of 3,000 kN · m, associated with a lateralload of 300 kN. Also note that the onset of the second line fit for the moment versus base rotation data points (3,430 kN · m) coin-cides with the onset of the third segment of the line fit for the lateralload versus lateral deflection data points (343 kN).

The maximum midspan deflections for the box girders in allsites were observed under the Array II configuration, which con-sisted of all trucks present in the exterior lane of the viaduct (Fig. 11). The magnitude for these deflections at sites PI, PV1,and VB were 37.0, 11.7, and 12.0 mm, respectively. These deflec-tions were obtained after taking into account the displacementsobserved at the supports as well as the local deflections at theflanges of the box girders. In all cases, the measured deflectionswere less than those allowed by the AASHTO specifications.

Prior to the static vehicular tests, preexisting cracks weremarked and identified for their monitoring during the tests. Themaximum width measured for preexisting flexural cracks at sitePI was 0.15 mm. The maximum crack widths measured before thetests at sites PV1 and VB were less than 0.03 mm. During the staticvehicular load tests, the maximum flexural crack widths measuredin site PI were 0.35 mm. The maximum crack widths at sites PV1and VB were 0.05 mm. All flexural cracks returned to their original

state once the tests were concluded. The measurement of crack widths was done using an Elcometer 143 crack width meter (Elc-ometer, Rochester Hills, MI). Inspection of the underside of thesuperstructure for cracks was possible with the use of Genie articu-lated and telescopic booms (Terex, Redmond, Washington).

The results of Array I and Array II of the static vehicular load testsat sites PV1 and VB showed that the distribution of vertical displace-ments in the cross-sections under study followed a linear trend(Fig. 12). This observation points to the conclusion that the cross-sec-tion displaced and rotated as a rigid body with respect to its support

point at the column bend. It might be also concluded that the mainsource of the vertical displacements of the cross section was the localdeformation of the column. The fact that the whole cross-section of the superstructure is able to rotate and displace as a rigid body sup-ports the idea that the connection between the column and the boxgirder had an adequate stiffness. From the designer ’s point of view,this would be the desired behavior for a posttensioned connection.

In sites PV1 and PV2, an approximate method to calculate thelateral displacement of the column under static vehicular load wasenvisioned assuming the column was a prismatic element with aninfinite axial stiffness and an equivalent fixed base estimated byadjusting the lateral displacements of a cantilevered column tothose measured during the tests. In the method, the demand asso-ciated with each of the arrays was represented by a concentrated

load and moment applied at the top of the column. The load cor-responded to the weight of the trucks and the moment was that associated with the eccentricity of the loads. The displacementscalculated with the approximate method showed good agreement with those measured during the tests.

Dynamic Vehicular Load Tests

The dynamic vehicular tests conducted at sites PI, PV1, and VBconsisted of the recording of acceleration and displacement data while trucks drove across the test site at different speeds. For thestudy of impact loading, a temporary speed bump was installed.The speed of the trucks varied from 10 to 73 km =h.

The maximum accelerations recorded for the dynamic vehicular

load tests were 40 and 668 m =s2 for the free and speed bump con-ditions, respectively. In order to estimate the dynamic amplificationfactors (DAF) with respect to the displacements for static vehicular load tests, the displacement records for the dynamic vehicular testswere first filtered using a low-pass filter to eliminate vibration com-ponents (Billing 1984), and then the equivalent static trace was ob-tained. Afterwards, the quotients of maximum amplitudes wereobtained for the filtered and unfiltered records.

Fig. 11. (Color) Arrays for the static vehicular load tests: (a) sites PI

and PV1; (b) site VB

Fig. 12. (Color) Vertical displacement of the superstructure measured during the static vehicular load tests: (a) site PV1; (b) site VB


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For the free drive tests of 490-kN and 686-kN loaded trucks,the highest DAFs at midspan were 1.60, 1.68, and 1.14 for sitesPI, PV1, and VB, respectively. The vertical deflections at midspanestimated with the former DAF were 7.8, 2.5, and 2.6 mm, respec-tively. For the speed bump condition in sites PI, PV1, and VB, thehighest DAFs at midspan were 1.70, 1.80, and 2.13, respectively.In this case, the vertical deflections associated were 26.9, 4.2, and2.8 mm, respectively. All deflections calculated using the obtainedDAF were less than the deflections measured during the staticvehicular load tests.

The largest values for the DAF were obtained for the speedbump condition and the run of the 160-kN and 196-kN unloadedtrucks. The values obtained for sites PI, PV1, and VB were 4.20,2.87, and 3.68, respectively. The corresponding deflections at midspan were 6.7, 0.4, and 1.1 mm. The range of DAF obtainedthroughout the dynamic vehicular load tests agrees well withvalues reported for bridges with similar characteristics (Paultreet al. 1992).

Ambient Vibration Tests in Sites PV1, PV2, and VB

Ambient vibration tests were conducted before and after any other tests in order to monitor any change in the original properties of the structure in all four sites. The ambient vibration tests focused

on determining the fundamental frequencies of the structure in thedirection perpendicular to traffic and the most relevant frequenciesof the girders.

The fundamental lateral frequency of the structure at site PV1was determined to be 1.12 Hz, which remained identical after thelateral load tests were concluded. In contrast, the first and secondnatural frequencies of the cantilevered column at site PV2 showeda slight reduction of around 3% from the before-collar to the after-collar tests. Indeed, as shown by the reduction in the lateral dis-placement of the column-footing assembly after the posttensionedcollar was cast, the presence of the collar increased the stiffnessesof the structural system. Furthermore, the development of flexuralcracks during the before-collar tests contributed to the change inthe natural frequencies of the assembly. The first two translational

frequencies for the column in the before-collar condition in sitePV2 were 1.12 and 1.51 Hz. For the after-collar condition, thesefrequencies changed to 1.10 and 1.47 Hz. In the case of site VB,the fundamental frequencies in the transverse direction of theviaduct were determined to be 5.9 and 1.4 Hz for the first stage(cantilevered column) and second stage (construction completed),respectively.

The natural frequencies determined by the ambient vibrationtests match very well those determined through the dynamic ve-hicular load tests at all test sites. Furthermore, the design firm pro-vided the authors with its calculations of the fundamental lateralfrequencies of the viaducts on the basis of their structural modeling.Table 3 summarizes the natural frequencies determined duringthese tests. Also, Table 3 shows the natural frequencies of the soil-structure ensemble determined on the basis of the structural modeldeveloped by the design firm for sites PV1 and VB. Data for other sites were not available. The largest differences between the mea-sured and calculated natural frequencies ranged from 7 to 24%. Thedifferences for the ratio of lateral/vertical frequencies ranged from 3to 11%. The natural frequencies calculated in the transverse direc-tion for sites PI, PV1, and PV2 were less than the correspondingdominant frequency of the soil (larger than 1.42 Hz). In contrast,the natural frequency estimated for the transverse direction for siteVB was larger than that for the soil at the site (around 1.08 Hz).

The fundamental frequency determined for the vertical compo-nent of the main span girder was only available for sites PV1

and VB (Table 3). The difference between calculated and observedvalues at these sites was 12 and 15%, respectively. The fundamentalmodal shapes in the lateral and vertical direction for all test sites areshown in Fig. 13. A good agreement between the values calculatedby the design firm and those obtained during the tests for sites PV1and VB may be observed in Fig. 13. The latter is also confirmed byvalues for the modal assurance criterion (MAC) (Allemang 2003).The value of the MAC calculated between analytical and experi-mental modal shapes at the central span for sites PV1 and VBwas larger than 0.95 for the first translational and vertical modes,and larger than 0.84 for the second vertical mode (asymmetric). Therange of vertical frequencies identified are similar to those reportedfor comparable structures (Paultre et al. 1992).

The equivalent viscous damping was determined through ambi-

ent vibration tests at sites PV1 and VB. The value associated withthe fundamental frequency of the main girder in the vertical direc-tion ranged from 2 to 6% of the critical damping for both sites. Theequivalent viscous damping associated with the lateral fundamentalfrequency ranged from 0.5 to 1.5%, and from 1 to 2% of the criticaldamping for sites PV1 and VB, respectively.

Analysis of Test Results

Effect of Foundation Stiffness at Site VB

The effect of foundation stiffness on the structural analysis of the viaducts was evaluated by comparing the measured natural

frequencies of the structure with those obtained by means of spe-cialized software and by means of the procedure indicated in thelocal design code (NTCDF 2004). The translational and rockingstiffnesses of the cantilevered column and its foundation at site VB,including SSI effects, were calculated using DYNA5 (Novak et al.1995) and NTCDF (2004). In the DYNA5 model, the soil profilewas defined and the group effect of the piles was taken in to ac-count. In contrast, the NTCDF code does not consider the groupeffect of piles. The stiffness of the footing-pile assembly was ob-tained as the sum of the stiffnesses of the footing and the pilegroup, which were calculated separately. Furthermore, the follow-ing two construction conditions were considered: (1) before thering of flowable fill was present, and (2) after the ring was cast.The stiffnesses were obtained for two frequency-dependent scenar-ios corresponding to the natural frequencies determined throughthe ambient vibration tests for the first (cantilevered column) andsecond stages (construction completed). A frequency-independent (static) scenario was also evaluated. The results of these analysesare shown in Table 4.

From Table 4, it is discernable that the differences between thecalculated stiffnesses for the static and dynamic cases are practi-cally negligible. Similar results for the translation and rocking stiff-nesses of a group of 2× 2 piles derived by Kaynia and Kausel(1982) using a rigorous theoretical model confirmed that there islittle difference between the static and the dynamic values com-puted. As for the effect of considering the presence of the ring

Table 3. Experimental and Analytical Natural Frequencies, in Hz

Data Direction Site PI Site PV1 Site PV2 Site VB

On-site tests Vertical (V) 3.42 4.00 3.61 4.65Lateral (L) 0.78 1.10 1.10 1.40Ratio (V/L) 4.4 3.6 3.3 3.3

Structuralmodel of thedesign firm

Vertical (V) N/A 3.51 N/A 3.95Lateral (L) 0.84 1.00 1.02 1.07Ratio (V/L) N/A 3.5 N/A 3.7


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of flowable fill, it increased the translational and rocking stiffnessan average of around 23 and 180%, respectively. The comparison

of the results in Table 4 with the estimated stiffnesses presented inTable 2 indicates that the ring of flowable fill should always beincluded in the analytical model.

Fig. 14 presents the results given in Table 4 in a plot, whichalso shows the variation of the translational (Kt) and rocking (Kr)stiffnesses calculated with DYNA5 and the NTCDF (2004) codeas a function of the frequency of the footing-pile ensemble for both the before-ring and after-ring of flowable fill conditions.Fig. 14 includes data points corresponding to the estimated trans-lational and rocking stiffness presented in Table 2. An ample rangeof frequencies (0–25 Hz) was considered in the analyses to better cover the whole dynamic response of the structure. The design firm addressed the soil-structure interaction using the procedure estab-lished by Zeevaert (1983). The equivalent translational and rocking

stiffnesses obtained were 268 MN=m and 11,788 MN · m =rad, re-spectively. Both values are significantly lower than those obtainedduring the tests and than those calculated using either DYNA5 or the NTCDF code with the flowable fill ring present.

In order to evaluate the general hypotheses assumed in the struc-tural analysis of the design firm, the model was used to evaluate theresponse of the cantilevered column at site VB. The model consid-ered the presence of a point load of 230 kN at the top of the can-tilevered column-footing ensemble. A second set of analyses wasrun for an effective horizontal load of 555 kN. In order to further understand the effect of the foundation stiffness on the response of the system, four values were considered: (1) design firm –calculated

values (ADesigner ); (2) DYNA5-calculated values for the before-ring of flowable fill condition (ADYNA5); (3) DYNA5-calculated

values for the after-ring of flowable fill condition (ADYNA5-R);and (4) estimated values derived from the ambient vibration tests(AExp). The values obtained using the simplified equations of theNTCDF (2004) code were very similar to those obtained for theADYNA5-R model; thus, they are not presented in this paper.

Fig. 15 shows the lateral displacement and rotation profilescalculated with the four values of foundation stiffness selected,

Fig. 13. (Color) Fundamental frequencies and relevant modal shapes of test sites

Table 4. Calculated Translational and Rocking Stiffnesses in Site VB

Analysismethod Stiffness

Before ring of flowable fill After ring of flowable fill

StaticFirst stage

(5.9 Hz)Second stage

(1.4 Hz) StaticFirst stage

(5.9 Hz)Second stage

(1.4 Hz)

DYNA5 Translational, MN=m 1,720 1,568 1,715 2,597 2,446 2,591Rocking, MN · m =rad 18,382 17,582 19,480 49,749 48,971 44,567

NTCDF Translational, MN=m 2,522 2,519 2,522 3,114 3,107 3,114Rocking, MN · m =rad 30,367 31,721 31,309 57,876 54,706 57,813

Fig. 14. (Color) Analytical impedance for the column-footing-piles

ensemble in site VB


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as well as the displacements and rotations measured during thefirst stage of the lateral load tests conducted at site VB. The modelthat best matched the measured displacements and rotations wasADYNA5-R. The largest difference between the results of the AExpand ADYNA5-R models was observed at the top of the column,

and was less than 12%.The differences between the ADesigner model and the models

based on DYNA5 values (ADYNA5 and ADYNA5-R) reflect the fact that the design firm used smaller SSI stiffnesses than those inferredfrom the lateral load tests. Furthermore, the four values for thefoundation stiffness were used in the full model developed by thedesign firm for the structural analysis. In this case, however, aneffective horizontal load of 431 kN was applied at the top of thecolumn. The results of the analysis using the values for the AExpcase indicated that 95% of the lateral load was resisted by just three columns as follows: 32% by column A, 43% by column B,and 20% by column C.

Fig. 16 shows the lateral displacements of columns A, B, and C,which were calculated with the full structural model and the four

foundation stiffnesses. Additionally, Fig. 16 shows the displace-ments measured during the second stage of the lateral load tests at site VB. Similarly to the cantilevered column model, the best matchfor the measured displacements was the model with A DYNA5-R val-ues. The differences between the results of the AExp and ADYNA5-Rmodels were less than 10%. Once again, the fact that the plots cor-responding to the ADesigner model lay further to the right of other models in the plots of Fig. 16 indicate that the hypotheses usedby the design firm resulted in conservative estimations. Due to

the fact that the spectral accelerations remain constant throughout a wide range of structural periods for the soft soil and the transitionzones in Mexico City, the design forces will remain constant, thusmaking the design more conservative when stiffnesses are under-estimated (NTCDF 2004).

Finally, the fundamental frequencies for the cantilevered columnand the full structural model were calculated. Table 5 summarizesthe results of the analysis. The average differences between thefrequencies calculated with the ADYNA5-R and AExp models andthose determined with the ambient vibration tests was 4 and 9%for the cantilevered column and full structural model, respectively.The average differences for the ADesigner and ADYNA5 models were19 and 24%, respectively.

Displacements and Frequencies for Vehicular Load Tests at Site VB

Using the full structural model of site VB, the lateral displacementsof the column were calculated for the vehicular load tests. Similarlyto the analysis for the effect of the foundation stiffness, four values(ADesigner , ADYNA5, ADYNA5-R, and AExp) for the SSI stiffness wereused and all five arrays for the load distribution were considered.Only the results of Array II, which showed the largest lateral dis-placements, are presented in the section.

The best prediction for the calculated lateral displacements of columns A, B, and C when subjected to the Array II condition wereobtained when the values of ADYNA5-R and AExp were used in thefull structural model. The average difference between the calculated

Fig. 15. (Color) Measured and calculated lateral displacements and rotations for lateral load tests of the cantilevered column in site VB:

(a) potentiometers; (b) inclinometers (I) and potentiometers (P)

Fig. 16. (Color) Measured and calculated lateral displacements for lateral load tests of the full structural model of site VB


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and measured displacements at the top of column B was under 9%for both ADYNA5-R and AExp. In contrast, the difference between thecalculated and measured displacements for the models using the

values of ADesigner and ADYNA5 were 36 and 21%, respectively.These differences are attributed to the fact that the SSI stiffnessesused by the design firm were conservatively smaller than thosedetermined through the tests.

The vertical displacements of the box girder calculated withthe full structural model resulted in similar values, regardless of theSSI stiffness used. The variation among the vertical displacementscalculated with the four stiffness values was less than 2%. The larg-est differences between the calculated and measured vertical dis-placements were under 20%, which were observed for the loadconditions of Array IV and V.

Lastly, the fundamental frequencies of vibration in the trans-verse and vertical directions of the viaduct were calculated usingthe full structural model and the four SSI stiffnesses. Table 6 sum-

marizes the transverse fundamental frequencies calculated for anunloaded condition and for Array III, which imposed a total loadof 1,249 kN and represents 64% of the service load used for thedesign of the superstructure.

The natural frequencies determined for the unloaded model us-ing the values for ADYNA5-R and AExp differ from those estimatedwith the dynamic vehicular load tests in less than 9%. In contrast,the frequencies calculated with the values for ADesigner and ADYNA5differ 24 and 16%, respectively, from the estimation based on the

dynamic vehicular load tests (Table 3). The lateral to vertical ratioof frequencies for the ADesigner and the AExp models is 3.7 and 3.2,respectively. The last value is close to the test-observed ratio of 3.3.The fundamental vibration frequencies in the vertical directionwere 3.95 and 4.14 Hz for the models with the values for ADesigner and AExp, respectively. These values have a difference of around15% with respect to those determined with the dynamic vehicular load tests (Table 3).

Comparison of Lateral Displacements

The lateral displacements for the lateral load tests and the staticvehicle load tests were compared for all sites. Fig. 13 shows theplots for the lateral load tests in sites PV1, PV2, and VB. Fig. 13shows the plots for Arrays I and II of the static vehicle load testsat sites PI, PV1, and VB.

The lateral displacement at the top of column B for site PV2was 1.8 times that corresponding to site PV1, which is comparablein terms of lateral load applied, height, and superstructure width(three traffic lanes). This might be attributed to the flexural crackingin the column, the cracking of the grout joint, the possibility of some voids in joint, the effect of the superstructure on the behavior of the column, or a combination of all four.

The deflected shape of column B in site VB showed a top dis-

placement that was 61 and 79% of the corresponding displacement in sites PV1 and PV2. The difference was due to a shorter column,thus, a smaller overturning moment (5,205 kN · m). The lateral dis-placements measured at the top of column B for Array I of the staticvehicle tests in site VB were 35% larger than those measured at the same height and under the same load condition for site PV1.A similar comparison for Array II of the static vehicle tests showedthat the displacements in sites PI and VB were 400 and 212%larger, respectively, than those at site PV1.

Furthermore, the asymmetric load distribution of Arrays I and II(Fig. 17) resulted in lateral displacements of the columns that weremuch larger than those measured during the lateral load tests. Staticvehicle load tests with Arrays I and II indicate that the box girder with the smallest displacements is the one in site PV1, which had

asymmetric flanges and three traffic lanes. In contrast, the super-structure with symmetric flanges and two traffic lanes at site PI wasthe most deformable.

The effect of increasing the number of trucks may be seen bycomparing the plots for Arrays I and II measured at sites PV1 andVB. The lateral displacement measured at the top of column Bfor the Array II load configuration in sites PV1 and VB was 1.1and 2.4 times, respectively, than that measured for Array I. Notethat Array II included four trucks in site PV1 and six trucks in

Table 5. Calculated Fundamental Frequencies for Site VB, in Hz

Foundationstiffness values

Cantileveredcolumn

Full structuralmodel

ADesigner 4.56 1.07ADYNA5 5.05 1.18ADYNA5-R 5.58 1.26AExp 5.75 1.28

Table 6. Calculated Natural Frequencies Determined with the FullStructural Model, in Hz

Foundationstiffness values Unloaded condition Array III

ADesigner 1.07 1.02ADYNA5 1.18 1.12ADYNA5-R 1.28 1.22AExp 1.27 1.21

Fig. 17. (Color) Lateral displacement measured at test sites: (a) lateral load tests; (b) static vehicle load tests


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site VB. Array I involved two trucks at both test sites. The incre-ment in lateral displacement suggests that at least four trucksshould be considered if the structural design is dominated by thelive load condition.

The initial flexural stiffness of column B was similar in sitesPV1, PV2, and VB. In contrast, the stiffness of column B in sitePI was around 6 times smaller. Indeed, the larger flexibility of column B in site PI resulted in much larger lateral displacementsthan those at other test sites. Furthermore, the circulation of loadedtrucks along the two-lane section in site PI induced larger vibra-

tions than those observed in the three-lane testing sites. The vibra-tions in site PI resulted in the oscillation of light posts to the point of generating discomfort and even a sense of insecurity in at least one of the truck drivers. It is suggested that the stiffness of columns supporting elevated viaducts be checked to avoid uncom-fortable vibrations.

Final Comments and Conclusions

The behavior exhibited by the elevated viaducts at all four test siteswas similar to that expected for a monolithic construction, despitebeing integrated by a series of prefabricated elements. The connec-tions and joints between the prefabricated elements were properly

detailed and constructed, showing a good performance in termsof both stiffness and strength for service loads during the tests.

The comparison of measured displacements at sites PV1 andPV2, which shared various structural characteristics, show that sim-ilar structures exhibited rather different responses. The tests at sitesPV1 and PV2 showed that the use of a posttensioned collar slabon top of the footing provided adequate rotational stiffness to thebase of the column. Nevertheless, the construction process wascomplicated and a simpler solution was suggested. The use of pre-fabricated column-footing ensembles in site VB was a result of the recommendation. Riobóo and Meli (2003), Haraldsson et al.(2013), and Osanai et al. (1996) have shown the structural perfor-mance limitations and construction difficulties associated withcandlestick-type columns. Their observations and the recommen-

dations of the authors led to a switch from a candlestick configu-ration to a prefabricated column-footing ensemble, which providedadequate structural continuity, better performance of the joint, andwas more efficient in terms of construction time. This structuralsolution is being used in the majority of elevated viaducts under design and construction in Mexico.

The tests at site VB indicate that the ring of flowable fill aroundthe footing contributed to increase the SSI stiffnesses. Nevertheless,the fill is yet another source for nonlinear response of the structureand its integrity is questionable under repetitive load cycles asso-ciated with seismic events. The use of SSI stiffnesses derived usinga model that included the ring resulted in calculated lateral dis-placements and rotations that were similar to those measuredduring the tests.

The results of the lateral load tests show that the frequenciesmeasured considering the SSI are dependent of the lateral load, thusindicating nonlinear effects between the foundation and the sur-rounding soil. The tests were able to determine the translationaland rocking stiffnesses of the system including the effects of SSI.

The vibration of the two-lane section of site PI during thedynamic vehicular load tests was larger than that observed at siteswith three traffic lanes. The vibration in site PI caused the light posts to oscillate and a sense of insecurity. The vibration at other test sites was barely noticed. A comparison of the response of thecolumns under lateral loads at the different sites indicates that, evenfor relatively similar structural designs, the observed behavior

might be rather different. Indeed, the absence of a code limitationfor the lateral displacements in elevated viaducts resulted in notice-ably different column stiffnesses, which conditioned a more flex-ible response of the two-lane test site with respect to the three-lanetest sites.

The authors emphasize that the maximum vehicular service loadused during the tests was greater than the one considered for design.This allowed for the study of the elevated expressways under anextreme service load scenario, even if improbable. Even under ex-treme loads, the lateral and vertical displacements measured were

under the limits allowed by design codes. Furthermore, the struc-tures recovered their original shape after the loads were removed.Under service conditions, lateral deformations under 0.04% areconsidered acceptable. Deformations around 0.1% are consideredinadequate, as users usually may perceive them as unsafe, whichwas experienced during the tests in site PI. It is recommended that future structures of similar characteristics be checked for reducedlateral displacements and minimal vibration to avoid uncomfortablesensations to drivers. It is also recommended that design codesclearly prescribe maximum lateral deformations that take into ac-count the stiffness of both the soil and the structure.

The fact that the observed response of the tested viaducts wasadequate for the type, magnitude, and distribution of loads appliedis a good indicator. However, this does not mean that the whole

viaduct would respond identically. For instance, the large differen-ces between the fundamental lateral frequencies measured and cal-culated for site VB may result in the subestimation of the seismicdesign forces, which may lead to an inadequate structural design.

The subestimation of the impedance functions for the SSI insite VB was the main cause for the large differences between theobserved and calculated value for the periods and the lateral dis-placements of the structure. The authors advocate the recommen-dations of Mylonakis and Gazetas (2000), BSSC (2001), Stewart et al. (2003), and Elnashai and Di Sarno (2008) with respect to themodeling of SSI. The main challenge is to develop a model that adequately represents the structure under service and ultimate loadconditions. The most reliable and accurate estimation of the effec-tive stiffnesses of the structural members under service and ulti-

mate load conditions is paramount in calculating the fundamentalfrequencies of vibration, which are directly linked to the estimationof design seismic forces and, thus, to the overall seismic perfor-mance of the structure.

The structural behavior displayed by the superstructure in alltest sites, particularly in terms of lateral stiffness, indicates that theposttensioned ensemble of prefabricated members may be consid-ered monolithic and continuous for service loads. Nevertheless, it would be wise to set up a monitoring scheme in order to detect anychanges of the structural properties during the service life of theviaducts, and to record their response under intense seismic events.The latter is of particular relevance for viaducts located in the soft soil region of the seismic zoning map of Mexico City in which thenonlinear effects associated with SSI may be significant. A fullstructural health monitoring system is to be deployed in site VBin the near future.

The data obtained during the field tests allowed for the assess-ment of the structural response of elevated viaducts in Mexico City,which has provided supporting evidence corroborating a number of analysis and design hypotheses used by the design firm. Allstructural elements monitored during the field tests showed an ad-equate behavior. Furthermore, a number of design changes aimedat expediting the construction without detriment to the safety of theproject have been developed since the first elevated viaduct wasbuilt in 2002. The field tests described in this paper have beeninstrumental in the design improvements that have taken place.


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One of the most important modifications addresses the design andconstruction of the footings. While they were built on site duringthe early projects, they have been prefabricated as an integral part of the column since 2008.

It is recommended that reversible-load field tests be carried out for cantilevered columns located in soft soils in order to evaluatethe potential degradation of the footing-soil interface. These testswould contribute to a better understanding of the viaducts responseunder cyclic loads, and are perfectly justified by considering thehundreds of similar columns that each elevated viaduct requires.

Acknowledgments

This research project was funded by the officials of Mexico Cityand of the State of Mexico. Thanks are extended to J.M. Riobóo, L.Cabrera, and A. Padilla for their comments and for providing thedesign firm models and corresponding structural and geotechnicaldocumentation. A special acknowledgement goes to J. Taboada,C. Hernández, E. Garcés, A. Borrego, C. Guerra, and M. Casasfor their support during the field tests and the provision of techni-cal information related to the projects. The authors also thank L.F.Fernández, M.A. Mendoza, G. Rodríguez, C. Cruz, J. Escobar,R. Mendoza, C. Terrones, R. Gómez, D. Aldama, J. Rodea,

V. Correa, C. Hernández, R. Mondragón, P. Trinidad, J. Rosales,L. Ibá ̃ nez, and a long list of undergraduate and graduate studentsfor their invaluable help during the field tests. The comments of R. Meli and S. Loera are well appreciated.

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ASCE - Mexico Viaduct Field Test

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